FINAL_GR07-14_Chickaloon_River_Bridge_Foundation_Report.pdf

Geotechnical Report No. 07-14

Glenn Highway Chickaloon River Bridge

FINAL FOUNDATION REPORT

AK DOT 135(2)

Matanuska-Sustina Borough, Alaska December 2014

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Contents

1. INTRODUCTION

1.1 Project Location

1.2 Scope of Project

2. PHYSICAL SETTING

2.1 Topography, Drainage and Vegetation

2.2 Existing Structure

2.3 Climate

3. SUBSURFACE CONDITIONS

3.1 Site Geology and Permafrost

3.2 Subsurface Investigation

3.3 Subsurface Profile

3.3.1 Fill

3.3.2 Alluvium

3.3.3 Bedrock

3.4 Groundwater Observations

3.5 Laboratory Materials Testing

4. FOUNDATION ANALYSIS AND RECOMMENDATIONS

4.1 Bridge Foundations Considerations

4.2 Scour

4.3 Seismic Hazard

4.4 Liquefaction

4.5 Bridge Layout

4.6 Design Loads

4.7 Foundation Design Methodology

4.8 Drilled Shaft Recommendations

4.9 Settlement

4.10 Lateral Analysis

5. DRILLED SHAFT CONSTRUCTION CONSIDERATIONS

6. ABUTMENT WINGWALLS

6.1 General Wall Recommendations

6.2 Soil Parameters for Retaining Walls

7. EXCAVATION AND EMBANKMENTS

8. LIMITATIONS

9. REFERENCES

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Appendices

APPENDIX A

Site Photographs

APPENDIX B

Boring Locations Map Fence Diagrams 1 through 4

APPENDIX C

Unified Soil Classification System Soil Descriptive Terminology Rock Descriptive Terms Boring logs and core photographs for CRB14-01 through CRB14-06

APPENDIX D

Laboratory Test Reports Soil Index Test Results Rock Uniaxial Compressive Strength Test Results Rock Point Load Index Test Results Rock Slake Durability Index Test Results

APPENDIX E

Final Design Loads Design Axial Geotechnical Resistance Charts

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Figure 1. Chickaloon River Bridge Project Location Map.

N

Chickaloon River Bridge

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1. INTRODUCTION

This report provides geotechnical investigation results, analysis, and foundation recommendations for a new bridge on the Glenn Highway over the Chickaloon River.

The project stationing referenced in this report is from the 95% Bridge Plan Set dated October 9, 2014.

1.1 Project Location

The Chickaloon River Bridge is located on the Glenn Highway about 30 miles northeast of Palmer, Matanuska-Susitna Borough, Alaska. The project location is shown on the maps in Figure 1 on page iii at the beginning of this report. The bridge is about 200 feet north of the intersection of the Glenn Highway with the Chickaloon River Road. The bridge crosses the Chickaloon River about 500 feet upstream of its confluence with the Matanuska River.

The coordinates for the new bridge are approximately latitude 61.7858 degrees north and

148.4538 degrees west.

1.2 Scope of Project

The existing bridge was built in 1956 and it is being replaced due to the poor condition of the existing structure. The replacement bridge will be constructed on a new alignment approximately 50 feet upstream (west) of the existing bridge within the Alaska Department of Transportation and Public Facilities (AKDOT&PF) right-of-way. The existing bridge will be used to carry public traffic during construction.

The total project length is 0.46 miles. The location of the new highway centerline will be on an interim alignment that ties into the existing alignment as a temporary condition until construction of the ultimate alignment along this section of the Glenn Highway. The existing alignment has a sharp curve to the south and a steep grade to the north. The proposed ultimate alignment improves the design speed around the sharp curve and reduces the steep grade but both of these improvements will require right-of-way acquisition. This project will utilize the interim alignment until the design and contract package for the ultimate alignment is completed and right-of-way acquisition completed.

The project will extend about 800 feet west of the bridge and about 900 feet east of the bridge.

This project will include cuts and fills up to about 20 feet in height.

The new bridge superstructure will have a cast-in-place concrete deck supported on prestressed concrete girders. The substructure will consist of drilled shaft foundations with pile caps at abutments and single column piers with hammerheads. The bridge will have a main span of 150 feet and two approach spans that are 100 feet each for a total length of 350 feet.

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2. PHYSICAL SETTING

2.1 Topography, Drainage and Vegetation

The Chickaloon River originates in the Talkeetna Mountains and flows generally south 35 miles into the Matanuska River. The drainage area is about 381 square miles (Frye, 2014). The project is located about 500 feet upstream of the confluence with the Matanuska River. The Matanuska River flows off the Matanuska Glacier located about 30 miles east of the project site and it is the main river flowing west between the Talkeetna and Chugach Mountain Ranges to the Cook Inlet and the Pacific Ocean.

Site photographs are attached in Appendix A. The site topography is mountainous terrain with relatively steep banks along both sides of the river below the existing bridge.

During site visits on May 9 and August 18, 2014, bedrock outcrops were observed below the existing and proposed bridge locations on the north river bank and just downstream of the existing bridge on the south river bank (see photos 8 through 14 in Appendix A).

Alluvial material observed in the river channel consists mostly of gravel, cobbles and boulders with some silt and sand deposits.

Upstream of the proposed bridge location the Chickaloon River is confined by steep rock outcrops on both sides and the terrain above the river is heavily vegetated (see the photo on the report cover and photo 9 in Appendix A). Downstream of the proposed bridge location the Chickaloon River floodplain broadens above the confluence with the Matanuska River with heavily vegetated banks (see photo 14 in Appendix A).

The water surface elevations and peak discharge levels reported in the Chickaloon River Bridge TS&L Hydraulics Report are summarized below for the proposed bridge configuration (Frye, 2014):

Recurrence Interval Water Surface Elevation (ft) Peak Discharge (cfs) 2-year 822.8 2934

100-year 826.2 8023 500-year 827.4 10,286

2.2 Existing Structure

The existing bridge is a continuous I-beam bridge approximately 250 feet long and 28 feet wide. The bridge is three spans with a 100-foot main span and two 75-foot approach spans. The abutments are reinforced concrete supported on driven 3-rail steel piles. The abutment walls are soldier pile tie-back walls with timber lagging (see photos 8 through 11 in Appendix A). The reinforced concrete piers each have two columns extending below the ground line. The plans for the existing bridge indicate the south pier is supported on driven 3-rail piles and the north pier is supported on spread footings.

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No overhead utilities were observed within the limits of the existing or proposed bridges.

2.3 Climate

The project is located in an area where the climate is influenced both by its proximity to the coast and by the surrounding mountain ranges. The site is located about 65 miles northeast of the Cook Inlet between the Talkeetna and Chugach Mountain Ranges. Table 1 summarizes average temperatures, precipitation, and snow depth measurements from the Sutton 2E weather station, located about 15 miles southeast of the site.

Table 1. Period of Record Monthly Climate Summary for Sutton 2E (508915).

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec Annual Average Max.

Temperature (F)

22.9 28.2 37.2 48.9 60.9 66.9 68 65.7 57.2 42.5 28.1 23.6 45.8

Average Min.

Temperature

(F)

9.1 11.5 17.3 25 32.4 40.7 46.3 44.1 35.9 25.9 14.8 10 26.1

Average Total Precipitation

(in.)

1.08 1.04 0.83 0.54 0.86 1.47 2.48 2.95 3.04 1.66 1.54 1.54 19.04

Average Total SnowFall (in.)

12.3 11 8.7 3.5 1 0 0 0 0.2 6.4 14 17.8 74.8

Average Snow Depth (in.)

16 18 16 5 0 0 0 0 0 1 6 13 6

Notes:

1. Period of Record : 1/ 1/1978 to 11/1/2011.

2. Percent of possible observations for period of record: Max. Temp.: 96.1% Min. Temp.: 96.2% Precipitation: 97.1%

Snowfall: 97.9% Snow Depth: 96.7%.

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3. SUBSURFACE CONDITIONS

3.1 Site Geology and Permafrost

The soil deposits at the project site consist of alluvial, colluvial, and glacial deposits overlying bedrock. Material interpreted as fill was encountered in borings off the edge of the existing roads.

The project limits are mapped as surficial deposits and Chickaloon Formation on the Geologic Map and Summary Geochronology of the Anchorage 1° x 3° Quadrangle (Winkler, 1992). The surficial deposits are described as widespread glacial, alluvial, colluvial, and lacustrine deposits, rock glaciers, and landslides that have not been subdivided on the geologic map. The Chickaloon Formation is described as carbonaceous mudstone, siltstone, conglomeratic sandstone, and polymictic conglomerate. The Chickaloon is more than 5000 feet (1500 meters) thick in the Matanuska Valley and the upper part of the formation contains numerous beds of bituminous coal.

On the Permafrost Map of Alaska (Ferrians, 1965) the site lies on the borderline of areas mapped as areas outside of permafrost region/generally free of permafrost and mountainous areas/generally underlain by isolated masses of permafrost. Permafrost was not encountered in any borings and it is interpreted that there is a low probability of encountering permafrost at this site.

3.2 Subsurface Investigation

Six subsurface investigation borings were completed for the proposed bridge foundations.

The boring locations are summarized on Table 2 and shown on the Boring Locations map in Appendix B. Fence diagrams summarizing the boring logs are contained in Appendix B.

Logs for all six borings (CRB14-01 through CRB14-06) are contained in Appendix C.

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Table 2. Summary of boring locations, elevations, and groundwater observations.

Boring No.

Station &

Offset1

Ground Surface

Elevation2 (feet)

Boring Depth (feet)

Water Level Observations2

Depth (feet)

Elevation (feet)

CRB14-01 403+88, 26.2’ Rt 848.4 79.7 31’ (WD) 817.4 (WD)

CRB14-02 403+94, 13.7’ Lt 845.9 79.5 3 3

CRB14-03 405+04, 3.8’ Rt 818.6 61.7 2.2’ (WD) 816.4 (WD)

CRB14-04 406+78, 3.9’ Lt 829.4 75.0

13.4’ (WD)

18.3’ (AC)

816.0 (WD)

811.1 (AC)

CRB14-05 407+72, 8.9’ Lt 858.0 125.5 40’ (WD) 818.0 (WD)

CRB14-06 407+87, 28.3’ Rt 868.1 87.0

29’ (WD)

55.7’ (AC)

839.1 (WD)

812.4 (AC)

NOTES:

1Boring locations and elevations determined by WFLHD Survey.

2WD = while drilling, AC = After completion and before casing withdrawal from borehole.

3No water level observations made.

The borings were completed between March 11 and April 12, 2014. A track-mounted CME 850 drill owned and operated by the AKDOT&PF was used to advance the borings.

Air rotary methods with a down-hole hammer and 4-inch casing were used to advance the hole. NQ-sized coring tools were used to retrieve samples of bedrock.

Standard Penetration Tests (SPTs) were completed at 2.5-foot to 5-foot intervals in overburden material. The Standard Penetration Test (SPT) consists of using a 140 pound hammer with a 30 inch drop to drive a two inch outside diameter split-barrel sampler 24 inches below the bottom of the bore hole and recording the number of blows for each six-inch increment of penetration. The middle two numbers (for the center 12 inches of penetration) are added together to determine the blows per foot (bpf), which is known as the N-value.

Borings were logged in the field by an AKDOT&PF Engineering Geologist in general accordance with the AKDOT&PF Alaska Geotechnical Procedures Manual (2007).

Samples were collected and shipped to Western Federal Lands Highway Division (WFLHD) in Vancouver, WA. The samples and draft boring logs were reviewed by the WFLHD Geotechnical Engineer and Engineering Geologist assigned to the project.

Selected samples of soil and rock were submitted to the WFLHD Materials Laboratory for testing.

3.3 Subsurface Profile

The borings encountered material interpreted as fill and alluvial deposits overlying bedrock. The thickness of the overburden (and depth to bedrock) varies significantly

GR07‐14 Page 6 across the site and the depth to bedrock is likely to vary by several feet within the limits of each bridge bent foundation. Fence diagrams summarizing the boring are contained in Appendix B and the boring logs and core photographs are contained in Appendix C.

3.3.1 Fill. Material interpreted as fill was encountered in Borings CRB14-01 and CRB14-02. These borings were drilled off the edge of the existing road near the proposed location for Abutment 1. The fill consists of silty clayey sand with gravel and silty sand; SPT N-values were between 34 and 38 blows per foot (bpf). SPT values corrected for drill hammer efficiency and overburden pressure are known as (N1)60 values. The (N1)60 values ranged from 70 to 78 bpf in the fill indicating the fill is very dense.

3.3.2 Alluvium. Alluvium consisting of silt- to boulder-sized material was encountered in all the borings at the ground surface or below the fill.

The alluvium recovered in the split barrel sampler consisted mostly of sand and gravel with varying amounts of silt and clay. Samples that were tested in the laboratory classified as silty clayey sand with gravel, silty clayey gravel with sand, silty sand with gravel, silty gravel with sand, and clayey sand with gravel.

Cobbles and boulders were encountered in some of the borings and it is interpreted that they are distributed throughout the alluvium across the entire project site. The material in the bottom of the river channel observed during site visits in May and August of 2014 consisted of silt to boulder-sized material (See photographs 8 through 14 in Appendix A).

Boulders observed were generally 1 to 2 feet in diameter, but some greater than 10 feet in diameter are present in the river channel and may also be present in the subsurface outside the current river channel.

Split barrel SPT samples in the alluvial deposits were collected in some of the borings.

The SPT values in the alluvial deposits ranged from 4 to 54 bpf and averaged 19 bpf.

The corrected SPT (N1)60 values ranged from 7 to more than 100 bpf and averaged 34 bpf, indicating the material was generally dense. It is likely that some of the SPT values were artificially high because of the coarse gravel- and cobble-sized particles encountered by the sampler in the alluvial deposits.

In CRB14-01 and CRB14-02 (near Abutment 1) the alluvium extended from approximately 21.5 to 26.5 feet below the ground surface. In CRB14-03 (near Pier 2) and CRB14-04 (near Pier 3) the alluvium extended 7.5 feet and 2 feet below the ground surface, respectively. In CRB14-05 and CRB14-06 (near Abutment 4) the alluvial deposits extended from approximately 7.5 to 12 feet below the ground surface.

The alluvial deposits are underlain by bedrock in all the borings. The bedrock contact elevation varies from boring to boring and is anticipated to vary within the foundation limits at each bent.

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3.3.3 Bedrock. Below the alluvium the borings encountered very weak (R0) to very strong (R5) sedimentary rock consisting of mudstone, sandstone, argillite, claystone, and siltstone with occasional coal layers. Mudstone was the predominant rock type encountered in all the borings. Layers of sandstone were encountered in all borings.

Argillite layers were only encountered in borings CRB14-01, CRB14-02, and CRB14-03 (near proposed Abutment 1 and Pier 2). Claystone layers were only encountered in CRB14-06 (near proposed Abutment 4). Siltstone layers were only encountered in CRB14-03, CRB14-04, and CRB14-05 (near proposed Pier 2, Pier 3, and Abutment 4, respectively).

Core recovery and rock quality designation (RQD) values in the rock were highly variable. Core recovery ranged from 0 to 100 percent in the borings and RQD ranged from 0 to 95 percent.

A total of 217 uniaxial compressive strength (UCS), diametral point load index tests, and axial point load index tests were completed. The UCS test and correlated point load index test results indicate that the intact rock strength is highly variable, both between borings and within each boring. The UCS and correlated point load UCS values ranged from 27 to 18,042 psi and averaged about 6200 psi.

Figure 1 shows the distribution of UCS values and correlated point load UCS values from all of the test data points.

Figure 1. Distribution of UCS and correlated point load UCS values.

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The UCS results were also highly variably within each rock type. A summary of the UCS test values and the correlated point load UCS values by rock type is presented in Table 3.

Table 3. Summary of UCS and Correlated Point Load UCS Values by Rock Type.

Rock Type # of Tests Average UCS

(psi) Minimum UCS

(psi) Maximum UCS

(psi)

ARGILLITE 9 6100 604 12,350

SANDSTONE 61 7700 27 17,255

CLAYSTONE 37 4700 78 13,766

MUDSTONE 101 5700 110 18,042

SILTSTONE 9 6500 3245 13960

ALL TESTS 217 6200 27 18,042

3.4 Groundwater Observations

Water level observations made in the subsurface investigation borings are summarized in Table 2 in Section 3.2. Groundwater was observed in the boring between elevations 811 and 840 feet.

The groundwater table is expected to be coincident with the river elevation. Perched water is likely to be present in permeable soil layers overlying bedrock.

Subsurface water is likely to be encountered during drilled shaft construction at or near the elevation of the river. Perched water is also likely to be encountered in the overburden material on top of bedrock, possibly at higher elevations than the river surface. Water elevations are likely to fluctuate with fluctuations of the Chickaloon River, particularly during spring thaw and following periods of heavy precipitation.

3.5 Laboratory Materials Testing

Selected split-barrel (SPT) soil samples were submitted to the WFLHD Materials Laboratory for routine index testing. Numerous rock core samples were submitted to the WFLHD Materials Laboratory for uniaxial compressive strength and point load index tests. Selected samples of rock core were also submitted to the Washington State DOT Materials Laboratory for slake durability tests. Laboratory test reports are contained in Appendix D.

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The following is a list of tests that were completed on the selected soil samples and the standards that were followed:

Particle Size Analysis – AASHTO T 88 Liquid Limit, Plastic Limit, and Plasticity Index – AASHTO T 89/90 Natural moisture – AASHTO T 265 Apparent Specific Gravity – AASHTO T 100

The results of the soil index tests were used to classify the samples according the Unified Soil Classification System (USCS). The USCS classification is reported on the boring logs.

The following is a list of tests that were completed on rock core samples and the standards that were followed:

Compressive Strength of Intact Rock Core – ASTM D 7012 (Method C) Point Load Strength Index – ASTM D 5731 Slake Durability Index – ASTM D 4644

Correlations between point load index and UCS were established following the procedures recommended in ASTM D 5731 by comparing the point load index values to UCS values within 1 foot elevation in the same boring. Separate linear correlation values were established for 1) mudstone and claystone, 2) siltstone, and 3) sandstone and argillite from the diametral test results. Only one correlation value was established for the axial point load test results and applied to sandstone, mudstone, and claystone because of the limited number of axial point load tests and because the linear relationship for all three rock types combined matched relatively well to the data points.

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4. FOUNDATION ANALYSIS AND RECOMMENDATIONS

The following analysis and recommendations reference the stationing and bridge dimensions in the 100% Bridge Plan set dated December 4, 2014. The bridge design is being completed by the Eastern Federal Lands Bridge Office in Sterling, Virginia.

4.1 Bridge Foundations Considerations

Site conditions, depth of bedrock, scour, site seismicity, design loading, cost, and constructability were the major factors considered when evaluating the foundation alternatives.

Shallow foundations were considered and dismissed as a foundation option due to the relatively high seismic loads that are anticipated, the potential for scour, and the relatively large excavations and associated impacts that would be required adjacent to the existing bridge.

Driven piling was considered at pier locations but later eliminated from consideration for the following reasons:

Piling within the limits of the flood plain would be more likely to catch debris, Large axial and lateral loading would require a larger pile group, resulting in a larger required footprint in/adjacent to the river riparian zone, Bedrock was encountered at depths less than 10 feet below the ground surface at pier locations. It would be difficult to drive piling to an adequate depth below the ground surface to satisfy scour and fixity requirements for a pile group.

Driven piling was considered at abutment locations but later eliminated from consideration for the following reasons:

Bedrock was encountered at depths less than 8 feet below existing ground surface near proposed Abutment 4. Based on the interpreted bedrock contact below the proposed abutment, one side of the proposed pile cap may rest on bedrock. We estimate that driven piling would not extend more than about 10 feet into bedrock before reaching refusal and/or damaging the pile. Thus it would not be possible to drive piling below the estimated scour depths. Also, due to the highly fractured nature of the rock mass it is desirable to extend foundations to greater depths.

Predrilling would therefore be required for the Abutment 4 piles. Predrilling would also likely be required at Abutment 1 in order to obtain adequate pile embedment into bedrock to obtain fixity and provide embedment below scour elevation.

Predrilling is estimated to have a cost similar to drilled shafts.

Drilled shafts can be constructed to extend well below the scour elevation and will be able to support relatively large axial and lateral loads when socketed into bedrock.

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Drilled shafts also require a minimal construction footprint in the river and can sometimes be constructed without a cofferdam. Use of drilled shafts has much less uncertainty in estimated tip elevations than driven piling.

Based on these considerations drilled shaft foundations are recommended for support of the new bridge.

4.2 Scour

Megan Frye, WFLHD Hydraulic Engineer, provided the following recommendation for scour in an email on July 22, 2014:

Pier scour is expected to be the primary scour mode. For the combination of contraction scour and pier scour, the 100-year and 500-year scour elevations are

808.5 feet and 807.8 feet respectively. For the combination of contraction scour and abutment scour, the 100-year and 500-year scour elevations are 815.1 feet and 815.0 feet respectively.

The scour analysis was completed assuming alluvial material. Bedrock was observed in the channel bottom and in the river bank on the north side of the river. Based on our field observations the bedrock observed is relatively resistant to scour compared to alluvial material. Also, the geotechnical borings indicate that the bedrock extends well above the 500-yr flood elevation near Abutment 4; therefore, scour was not considered for the Abutment 4 geotechnical analysis.

The recommended scour elevations listed above were used at the Abutment 1, Pier 2, and Pier 3 locations.

4.3 Seismic Hazard

The General Procedure outlined in Section 3.10 of the AASHTO LRFD Bridge Design Specifications (2014) was followed to characterize the seismic hazard. The General Procedure uses mapped gridded values of peak ground acceleration, 0.2 second spectral acceleration, and 1.0 second spectral acceleration to develop the 5-percent-damped-design response spectrum chart.

These procedures use a design earthquake with a return period of 1000 years, or a 7 percent probability of exceedance (PE) in 75 years. The AASHTO Earthquake Ground Motion Parameters program (version 2.10) was used to determine the peak ground acceleration and spectral accelerations. The site class was selected after correlating published compression wave (p-wave) velocities (Pintner, et al, 2007) to shear wave velocities. Site factors were selected from Table 3.10.3.2-1 through 3.10.3.2-3 in AASHTO (2014) and multiplied by the mapped peak ground accelerations and spectral

GR07‐14 Page 12 accelerations in order to determine the modified peak ground acceleration and spectral accelerations. The results of the hazard analysis are presented in Table 4.

Table 4. Summary of Seismic Parameters (7% PE in 75 years).

Site Class (by Soil Type and Profile) C

Horizontal Peak Ground Acceleration (PGA) 0.54g

Horizontal Response Spectral Acceleration at Period of 0.2 sec (Ss) 1.22g

Horizontal Response Spectral Acceleration at Period of 1.0 sec (S1) 0.47g

Site Factor for Zero-Period on Acceleration Spectrum (Fpga) 1.00

Site Factor for Short-Period Spectral Acceleration (Fa) 1.00

Site Factor for Long-Period Spectral Acceleration (Fv) 1.33

Modified Horizontal Peak Ground Acceleration (As) 0.54g

Modified Horizontal Response Spectral Acceleration at Period of 0.2 sec (SDS) 1.22g

Modified Horizontal Response Spectral Acceleration at Period of 1.0 sec (SD1) 0.63g

Seismic Zone 4

The modified peak ground acceleration and spectral acceleration values in Table 4 were used to develop the 5-percent-damped-design response spectrum shown in Figure 2 below.

Figure 2. 5-Percent-Damped-Design Response Spectrum.

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4.4 Liquefaction

The soils encountered at this site are not susceptible to liquefaction. The cohesionless soil layers encountered above bedrock were not saturated and therefore not susceptible to liquefaction.

4.5 Bridge Layout

The new bridge structure will have the following characteristics:

Three spans Offset approximately 50 feet west of existing bridge Total length of 350 feet Main span length of 150 feet Approach span lengths of 100 feet Total width of 45 feet Approximate skew angle of 0 degrees

At each abutment the superstructure will be supported on two drilled shafts and pile caps.

At the pier locations the superstructure will be supported by a single drilled shaft and pier column with a hammerhead.

4.6 Design Loads

The Eastern Federal Lands Bridge Office provided final axial compressive loads for the abutments and piers on September 29 and final lateral loads for the piers on October 2, 2014, via email. The final loads are attached in Appendix E.

The following loads were the highest calculated axial loads for each limit state:

Abutment 1 Maximum Loads Strength Limit State I axial compressive load = 1274 kips / shaft Service Limit State II axial compressive load = 991 kips / shaft Extreme Event Limit State I lateral shear load = 990 kips / shaft (transverse to bridge)

Pier 2 Maximum Loads Strength Limit State I axial load = 5542 kips / pier Service Limit State II axial load = 4307 kips / pier Extreme Event Limit State I lateral shear load = 1267 kips / pier (longitudinal to bridge) Extreme Event Limit State I moment at base of Pier = 40,800 kip-ft

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Pier 3 Maximum Loads Strength Limit State I axial load = 5486 kips / pier Service Limit State II axial load = 4262 kips / pier Extreme Event Limit State I lateral shear load = 1350 kips / pier (longitudinal to bridge) Extreme Event Limit State I moment at base of Pier = 40,800 kip-ft

Abutment 4 Maximum Loads Strength Limit State I axial compressive load = 1184 kips / shaft Service Limit State II axial compressive load = 921 kips / shaft Extreme Event Limit State I lateral shear load = 990 kips / shaft (transverse to bridge)

The abutment loads include the dead load to the bottom of the pile cap, but do not include the dead load of the drilled shaft above the design scour elevation. The maximum dead load of the drilled shaft section above scour elevation was calculated to be 103 kips for each 5-foot diameter drilled shaft at Abutment 1 and 76 kips for each 5-foot diameter drilled shaft at Abutment 4.

The pier loads include the dead load of the columns and hammerhead above the top of the drilled shaft, but do not include the dead load of the drilled shaft above the design scour elevation. The maximum dead load of the drilled shaft section above scour elevation was calculated to be 107 kips for a 10-foot diameter drilled shaft at pier locations.

Downdrag loads due to settlement are not anticipated.

4.7 Foundation Design Methodology

Drilled shaft foundations were designed in accordance with the 7th Edition of the AASHTO LRFD Bridge Design Specifications (2014). The 7th Edition includes updates to the geotechnical design of drilled shaft foundations in rock that are consistent with the recommended design procedures in the FHWA/NHI Drilled Shaft Design Manual (Brown, et al, 2010). The main difference between the new design methodology and previous editions of AASHTO is that the new procedures utilize geologic strength index (GSI) to characterize the strength and modulus of the rock mass rather than rock mass rating (RMR).

Drilled shafts were designed using a combination of skin resistance and tip resistance.

Side resistance was neglected within the zone of permanent casing. Resistance factors for single drilled shafts at Pier locations were reduced by 20 percent in accordance with

AASHTO (2014).

Both axial compressive and uplift resistance were evaluated at the abutments. Since single shafts are being used at Pier 2 and Pier 3 only axial compressive resistance was evaluated at those locations. Charts of axial resistance versus depth are provided in Appendix E.

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4.8 Drilled Shaft Recommendations

Recommendations for drilled shafts are summarized in Table 5.

In addition to the design loads, subsurface conditions and settlement criteria were considered when establishing the recommended drilled shaft tip elevations.

As discussed in Subsection 3.3, the bedrock elevation varies across the site and it is expected to vary within the limits of each foundation. The drilled shafts should be extended into bedrock to the tip elevations listed in Table 5.

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Table 5. Drilled shaft recommendations.

Abutment 1 Pier 2 Pier 3 Abutment 4

Station at Centerline of Bent 404+09.76 405+07.84 406+57.84 407+55.92

Top of Drilled Shaft Elevation

848.64 LT/

847.57 RT

819.50 825.70

858.08 LT/

857.29 RT

Number of Drilled Shafts per Bent 2 1 1 2

Approximate Top of Bedrock Elevation at Road Centerline1

824+/- 811+/- 826+/- 853+/-

Permanent Casing Inside Diameter (feet)

5 10 10 5

Permanent Casing Wall Thickness (inches)

0.75 0.75 0.75 0.75

Permanent Casing Top Elevation

848.37 LT/

847.30 RT

819.50 825.70

857.81 LT/

857.02 RT

Permanent Casing Tip Elevation 811.40 797.00 812.00 825.02

Length of Permanent Casing (feet)

36.97 LT/

35.90 RT

22.5 13.7

32.79 LT/

32.00 RT

Minimum Drilled Shaft Socket Diameter (feet)

4.5 9.5 9.5 4.5

Recommended Drilled Shaft Tip Elevation

800.40 777.00 788.00 814.02

Minimum Length of Socket1,2 (feet)

11.0 20.0 24.0 11.0

Total Length of Drilled Shaft3

(feet)

48.24 LT/

47.17 RT

42.5 37.7

44.06 LT/

43.27 RT

Geotechnical Resistance Factors - Side resistance / tip resistance

0.55/ 0.50

0.44/ 0.40

0.44/ 0.40

0.55/ 0.50

Nominal Axial Compressive Resistance (kips)

3900 18,900 14,450 3350

Factored Strength Limit State Axial Compressive Resistance for

Single Drilled Shaft (kips) 2000 7800 6050 1750

NOTES:

1 The bedrock elevation varies across the site and it is expected to vary within the limits of each foundation.

2 Socket lengths measured from permanent casing tip elevation to recommended drilled shaft tip elevation.

3Total length measured from top of drilled shaft elevation to recommended drilled shaft tip elevation.

GR07‐14 Page 17

4.9 Settlement

The estimated settlement of the drilled shafts due to the service limit state loads listed in Subsection 4.6 is less than ½ inch at the top of the drilled shaft. Load-settlement calculations for the loads listed in Table 5 indicate that most of the load will be transferred into the rock mass through side resistance and only a small portion of the load will be transferred to the base. The loads in Section 4.6 do not exceed the limit that would cause side shear failure, which would transfer a more substantial portion of the load to the base, reduce the side resistance, and result in larger deformations at the top of the shafts.

4.10 Lateral Analysis

Table 6 provides the soil and rock parameters that are recommended to analyze lateral foundation loads. The values in Table 6 do not include factors of safety or load factors.

A resistance factor of 1.0 is specified in AASHTO (2014) for horizontal geotechnical resistance of a single shaft or shaft group.

Due to the highly variable strength of bedrock encountered in the borings a sensitivity analysis was completed in order to assess the effects of different soil and rock strength properties on the results of the LPile analysis. The results of the analysis (shear and bending moment diagrams and plot of deflection versus depth) did not change significantly when a range of rock strength parameters were input. The values in Table 6 are average values for the layers.

GR07‐14 Page 18

Table 6. Recommended soil parameters for lateral analysis.

Properties for

Sand Properties for Strong Rock and

Weak Rock Elev.

(feet)

LPILE

Material

(pcf)

(pcf)

(deg) k (lb/in3) qu

(psi) Erm (psi)

RQD

(%) krm

Abutment 1 (use effective unit weight below elevation 826)

> 823 Sand

(Reese) 115 53 39 90 - - -

804-823 Weak Rock

160 98 - - 500 5000 0 0.0001

< 804 Strong Rock

165 103 - - 5000 - - -

Pier 2 (use effective unit weight below elevation 826)

>811 Sand

(Reese) 125 63 43 125 -

797-811 Weak Rock

160 98 - - 500 5000 0 0.0001

787-797 Strong Rock

165 103 - - 4200 - - -

777-787 Strong Rock

165 103 - - 2400 - - -

<777 Strong Rock

165 103 - - 5000 - - -

Pier 3 (use effective unit weight below elevation 826)

>814 Weak Rock

160 98 - - 1500 13000 0 0.0001

<814 Strong Rock

165 103 - - 4400 - - -

Abutment 4 (use effective unit weight below elevation 826)

>844 Sand

(Reese) 125 63 37 25 -

827-844 Weak Rock

165 103 - - 1500 50,000 25 0.0001

<827 Strong Rock

165 103 - - 4000 - - -

NOTES:

1. = moist unit weight.

2’ = effective unit weight.

3’ = effective internal angle of friction.

4. k is the constant giving the variation of soil modulus with depth in the equation Es = kz, where Es = soil modulus, and z = depth below the ground surface.

5. qu = unconfined compressive strength.

6. Erm = elastic modulus of rock mass.

7. RQD = rock quality designation.

8. krm = rock strain factor.

GR07‐14 Page 19

5. DRILLED SHAFT CONSTRUCTION CONSIDERATIONS

Significant issues related to drilled shaft construction at this site include, but are not limited to, fluctuations in water levels of the Chickaloon River, presence of boulders in overburden soils, variability in rock strength, variations in elevation of the bedrock surface, and hole stability.

Equipment will need to access the pier locations to install the drilled shafts and construct the piers. Fluctuation in the river elevation may impact access to the river. The top of drilled shaft elevations at Pier 2 and Pier 3 are 819.5 and 825.7 feet, respectively.

Groundwater levels encountered in the borings at the time of drilling ranged from 811 to 818 feet. River elevations for the 2-year and 100-year recurrence events are 822.7 and

825.9 feet, respectively.

Based on the results of the geotechnical borings and surface observations of the Chickaloon River channel and adjacent flood plain it is anticipated that boulders up to 10 feet in diameter may be encountered during the drilling and installation of drilled shaft foundations. The contractor should be prepared to remove boulders that impede or prevent advancement of the shaft.

The thickness of the overburden (and depth to bedrock) varies significantly across the site and the depth to bedrock is likely to vary by several feet within the limits of each bridge bent foundation. Fence diagrams summarizing the boring and showing the relative elvations of the layers encountered in each boring are contained in Appendix B. Drilling equipment may wander out of vertical tolerance when sloping bedrock is encountered and appropriate measures must be taken to maintain a plumb drill hole.

The overburden soils encountered in the borings are likely to slough below the water table and possibly above the water table if excavated with an open hole. Wet methods of drilled shaft construction will likely be required to protect the side walls of the drilled shaft excavation through the overburden soils during construction.

The bedrock encountered in the borings is highly fractured and the contractor should be prepared to encounter water in the drilled shaft excavation even if temporary casing is socketed into the bedrock. Slake durability index test were completed on samples of core from the borings to determine if the rock will weather and degrade rapidly. The test results are contained in Appendix D. Slake durability index values ranged from 83.5% to 98.1%. Slake durability index values less than 60% indicate that bedrock exposed in drilled shaft bore hole walls may be susceptible to rapid degradation when water is introduced during drilled shaft excavation (Brown et al, 2010).

The drilled shaft design includes both skin friction along the sides of the drilled shaft socket and tip resistance at the base of the socket. Therefore, any temporary casing must be removed from the completed drilled shaft socket. If an oversized hole is drilled for

GR07‐14 Page 20 installation of permanent casing the void between the outside of the permanent casing and the soil/rock must be sealed by tremie grouting or other approved methods. The bottom of the shaft must be cleaned and a shaft inspection device (SID) is recommended to verify that any remaining sediment on the bottom of the excavation is within specified limits.

The bedrock at the site is highly variable and unconfined compressive strength (UCS) test results and point load test results correlated to UCS tests on rock core specimens ranged from as low as 27 psi (extremely weak rock[R0]) to as high as 18,042 psi (very strong rock [R5]). The UCS values were generally between 4700 and 7700 psi (medium strong [R3] to strong [R4] rock). Uniaxial compressive test results are discussed in Subsection

3.3.3 and contained in Appendix D. Rock coring methods will likely be required in order to penetrate the shafts to the required depths.

Cross-hole Sonic Logging (CSL) should be completed on each drilled shaft to insure the integrity of the shaft. Each rebar cage should include one CSL access tube per foot diameter of drilled shaft (i.e., at least 10 CSL access tubes for a 10-foot diameter drilled shaft). The access tubes should be spaced evenly on the rebar cage and securely fastened to either the inside or outside of the cage.

Concrete poured in the drilled shafts should have a minimum 28-day compressive strength of 4000 psi.

The WFLHD Geotechnical Team should be part of the reviewing effort of the contractor’s drilled shaft submittals to evaluate shaft design and construction appropriateness based on the subsurface conditions. A member of the Geotechnical Team should also be present when excavation of the drilled shafts begins to ensure that the subsurface conditions are consistent with the materials encountered in the subsurface exploration.

GR07‐14 Page 21

6. ABUTMENT WINGWALLS

The wingwall on the north (left) side of Abutment 1 will be a reinforced concrete cantilever semi-gravity wall separated from the abutment with a joint. The wingwall on the south (right) side of Abutment 1 will be monolithic with the abutment. Both of the wingwalls on Abutment 4 will be monolithic with the abutment.

6.1 General Wall Recommendations

The following geotechnical recommendations apply to the retaining walls:

Use non-frost susceptible (NFS) structural backfill behind wing walls. NFS is defined as minus 3-inch material with less than 5 percent by weight passing a No. 200 Sieve.

Remove topsoil and vegetation in sections of the walls bearing on native ground and compact the subgrade prior to constructing the wall footings. Soft or unsuitable soils are not anticipated, but if they are encountered, they should be over-excavated and replaced with compacted Structural Backfill

Install a perforated corrugated polyethylene drain behind walls to ensure adequate drainage. Provide multiple weep holes for drainage through the wall and extend the drainage pipe to daylight out to the limits of the fill in front of the wall.

Wall footings not supported on drilled shafts should bear a minimum of 6 feet below the ground elevation on the exterior side of the wall for frost protection.

Construct a flat 4-foot wide bench (with minimal slope to drain away from the wall face) along the wall face.

Fill slopes in front of the walls should be constructed at 1V:2H or flatter.

Aggressive erosion protection and revegetation measures should be taken to assure that these slopes in front of the walls are quickly revegetated. Riprap outlet protection should be placed in areas below drains and downspouts.

6.2 Soil Parameters for Retaining Walls

The soil parameters listed in Table 7 may be used to design the wing walls. The parameters listed in Table 7 are for vertical walls with level ground at the top of the wall.

GR07‐14 Page 22

Table 7. Recommended Soil Parameters for Abutment Wingwall Design.

Parameter Design Value*

Moist Unit Weight 135 pcf

Angle of Internal Friction 36 degrees

Active Earth Pressure Coefficient, Ka 0.25

At-Rest Earth Pressure Coefficient, K0 0.41

Passive Earth Pressure Coefficient, Kp 8.2

Wall-Backfill Interface Friction Angle, 22 degrees

*Design values determined in accordance with Chapter 3 of AASHTO LRFD (2014) based on a drained condition behind the wall.

The above values do not include factors of safety or load factors. Appropriate load and resistance factors should be used with these parameters. It is generally assumed that compacted backfill creates at-rest lateral earth pressure behind unyielding walls and that a slight movement at the top of the wall (0.002 x wall height) creates active lateral earth pressures behind walls. When passive lateral earth pressures are applicable, wall movements of 0.02 x wall height at the top of the wall are necessary to reach the earth pressures listed above.

The procedures outlined in Appendix A11 of AASHT LRFD should be used to determine the seismic forces exerted on the walls.

GR07‐14 Page 23

7. EXCAVATION AND EMBANKMENTS

The project will extend about 800 feet west of the bridge and about 900 feet east of the bridge. This project will include cuts and fills up to about 20 feet in height.

Roadway embankment will likely be constructed with material from nearby cuts. The only substantial cut on the project is the cut near proposed Abutment 4. Material excavated near Abutment 4 will consist mostly of silty gravel, sand, mudstone, and sandstone bedrock which is adequate for use as embankment fill (see Boring Logs CRB14-05 and CRB14-06 in Appendix C and the accompanying laboratory test reports in Appendix D).

If layers of coal are encountered in the excavation it should not be used as embankment material.

Embankments constructed with material excavated on site can safely be constructed as steep as 1V:2H.

GR07‐14 Page 24

8. LIMITATIONS

This report was prepared to provide subsurface investigation results and final geotechnical recommendations for a new bridge over Chickaloon River. Professional judgments and recommendations are presented in this report. They are based partly on evaluation of the technical information gathered, partly on historical reports, and partly on the Geotechnical Team’s general experience with subsurface conditions in the area. It should be noted the borings might not represent potentially unfavorable subsurface conditions between borings. The observed water levels and/or conditions indicated on the boring logs are as recorded at the time of the exploration. The water levels and/or conditions may vary considerably, with time, according to the climate, rainfall, and other factors and are otherwise dependent on the duration of, and methods used in, and interpretations made from the exploration program.

If, during construction, soil conditions are encountered that vary from those discussed in this report or historical reports, or if design loads, project stationing, and/or configurations change, the WFLHD Geotechnical Team should be notified immediately in order that it may evaluate effects, if any, on the recommendations contained in this report. The recommendations presented in this report are applicable only to this specific site. These data are not to be used for other purposes.

Questions regarding this report should be addressed to Brian Collins, WFLHD Senior Geotechnical Engineer, at (360) 619-7657, or brian.collins@dot.gov.

GR07‐14 Page 26

9. REFERENCES

AASHTO LRFD Bridge Design Specifications, 7th Edition (2014), American Association of State Highway and Transportation Officials, Washington, D.C.

Alaska Geotechnical Procedures Manual (2007) Alaska Department of Transportation and Public Facilities.

Brown, D.A., Turner, J.P., Castelli, R.J. (2010) Drilled Shafts: Construction Procedures and LRFD Design Methods, FHWA-NHI-10-016, Geotechnical Engineering Circular No.

10, National Highway Institute, U.S. Department of Transportation, Federal Highway Administration, Washington, D.C.

Frye, M. (2014) Chickaloon River Bridge TS&L Hydraulics Report, US DOT, Federal Highway Administration, Western Federal Lands Highway Division.

Isenhower, W.M., and Wang, S-T., (2011) Technical Manual for LPile, Version 6, Ensoft, Inc.

Pintner, R.M., French, R.B., and Riddle, C.H. (2007) Slope Stabilization Evaluation Seismic Refraction Survey, Glenn Highway, MP 92 to 97, Cascade to Hicks Creek, Project No. NH-0A1-5(17)/52507, R&M Consultants, Inc., Anchorage, Alaska, Prepared for Alaska Department of Transportation and Public Facilities.

Western Regional Climate Center, Cooperative Climatological Data Summaries, Sutton 2 E, Alaska (508915) www.wrcc.dri.edu, accessed September 24, 2014.

Winkler, G.R. (1992) Geologic Map and Summary Geochronology of the Anchorage 1° x 3° Quadrangle, Southern Alaska, USGS Miscellaneous Investigations Series, Map I-2283.

APPENDIX A

Site Photographs

Appendix A

Page A‐1

1. Looking northeast from about Station 397+50, 09May2014.

2. Looking north from about Station 400+00,09May2014.

Page A‐2

3. Looking north from about Station 401+00, 09May2014.

4. Looking north from about Station 401+50, 09May2014.

Page A‐3

5. Looking west from about Station 404+00, 09May2014.

6. Looking southwest from about Station 404+00, 09May2014.

Page A‐4

7. Looking north from about Station 404+00, 09May2014.

8. Looking northwest from about Station 404+50, 09May2014.

Page A‐5

9. Looking northwest from about Station 405+50, 09May2014.

10. Looking north from about Station 405+50, 09May2014.

Page A‐6

11. Looking southeast from about Station 405+50, 09May2014.

12. Looking west from about Station 406+50, 18Aug2014.

Page A‐7

13. Looking south from about Station 407+50, 09May2014.

14. Looking south from about Station 407+50, 09May2014.

Page A‐8

15. Looking west from about Station 408+50, 09May2014.

16. Looking southwest from about Station 409+00, 09May2014.

APPENDIX B

Boring Locations Map

Fence Diagrams 1 through 4

8 7

M P

N

CRB14-01

CRB14-02

CRB14-06

CRB14-05

CRB14-04

CRB14-03

Line to be constructed existing pavement

Edge of

Existing bridge

Proposed bridge

Edge of water

50 10002550

SCALE IN FEET

pier (typ.)

Existing bridge bridge abutment (typ.)

Existing

X.1

STATE PROJECT

NUMBER

SHEET

AK

DOT 135(2)

A

M

S e p te m b e r

U

S S u r ft

D c m y fi le s p w p r o d u c ti o n d a k s a .d g n

C h e c k e d b y

D e s ig n e d b y

BORING LOCATIONS

CHICKALOON RIVER BRIDGE

GLENN HIGHWAY

D A

O R

R E

VI

R

N O

O L

A K

CI

H

C drilled shaft (typ.)

Proposed A b u tm e n t

A b u tm e n t

P ie r

P ie r

Appendix B

Page B-1

0 50 100 150 200 250 300 350 400

0 50 100 150 200 250 300 350 400

-15 -10 -5 0 5 10

-15 -10 -5 0 5 10

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD 100% 0

SC-SM

SC-SM

GC-GM

GP-GM

SP

ML

SP

SW

MUDSTONE

COAL

MUDSTONE

COAL

MUDSTONE

SANDSTONE

MUDSTONE

ARGILLITE

MUDSTONE

SANDSTONE

4.5'

9.5'

14'

19.5' 21.5' 22.5' 23.5' 26.5'

32.5'

35.5' 37' 38'

49.5' 51.5'

70' 72'

76'

79.7'

PL

CRB14-01

LL

N Value(N1)60

403+88,26.15' RT. EL: 848.4'

(N1)60 50

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD 100% 0

SM

GM

BOULDERS

GP

SM

MUDSTONE

MUDSTONE

COAL

COAL

ARGILLITE

MUDSTONE

SANDSTONE

COAL

MUDSTONE

SANDSTONE

5.5'

8.5' 11' 12'

21.5'

25'

36'

41.4'

45.5'

54.5'

62' 64'

65.5'

79.5'

CRB14-02N Value(N1)60

403+94,13.7' LT. EL: 845.9'

(N1)60 50

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD 100% 0

GW

MUDSTONE

COAL

MUDSTONE

SILTSTONE

ARGILLITE

SANDSTONE

MUDSTONE

SANDSTONE

MUDSTONE

SANDSTONE

MUDSTONE

7.5'

19.5' 21.5'

28'

32'

42' 42.5' 45.5'

47'

51'

61' 62'

CRB14-03N Value(N1)60

405+04,3.78' RT. EL: 818.6'

(N1)60 50

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD 100% 0

GM

SP

SANDSTONE

COAL

MUDSTONE

COAL

SILTSTONE

SANDSTONE

SILTSTONE

COAL

COAL

SANDSTONE

MUDSTONE

SANDSTONE

MUDSTONE

SANDSTONE

MUDSTONE

2' 3.5'

10' 12.5'

14' 15'

15.5'

22.5' 23.2'

25'

35'

40'

56.7' 58.7'

60'

64'

74.25'

CRB14-04N Value(N1)60

406+78,3.86' LT. EL: 829.4'

(N1)60 50

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD 100% 0

GM

SP

GM

GP

COAL

SANDSTONE

COAL

MUDSTONE

SANDSTONE

MUDSTONE

COAL

MUDSTONE

COAL

MUDSTONE

SANDSTONE

MUDSTONE

SILTSTONE

COAL

COAL

MUDSTONE

COAL

SANDSTONE

SANDSTONE

SANDSTONE

MUDSTONE

SANDSTONE

MUDSTONE

3.5' 5.5' 7.5'

12' 14'

19' 20'

40' 41'

55'

60.5'

71.5' 74' 76'

80.5'

84' 85.5' 87.7'

90.5'

101.5' 103.5'

123.8' 124.3' 125.5'

CRB14-05N Value(N1)60

407+72,8.89' LT. EL: 858.0'

(N1)60 50

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD

RQD 100% 0

GM

GW

MLS

MUDSTONE

SHALE

SANDSTONE

MUDSTONE

COAL

SANDSTONE

SHALE

SANDSTONE

SANDSTONE

4.5' 6.5' 7.5'

44'

51.5' 53'

61' 63.5'

73.5'

77.5'

86' 87'

CRB14-06N Value(N1)60

407+87,28.33' RT. EL: 868.1'

(N1)60 50

Alaska

Chickaloon River Bridge AK DOT 135(2)

5 ft intervals

RQD (%)

RQD

PLATE: 1

-Longitude ° Depth

Position -26.15000002

40397.89858044 Right, Front

405+04,3.78' RT

407+87,28.33' RT

(N1)60 Value

79.7

868.161.786256 -148.453317

79.5

Recovery (%)

61.785983 61.786234

-148.454072 -148.45385

-148.453777

DATE: 16 Oct 2014

CRB14-05

CRB14-06

CRB14-01

CRB14-03

858.0

Elev.

Ending

848.4 61.785279 61.785559

845.9 818.6 829.4

SUBSURFACE FENCE DIAGRAM

Fence 1

CRB14-01 through CRB14-06

At completion After drilling

WATER LEVELS

While drilling

CRB14-04

-148.453571

<--- Scale (%)

Recovery and RQD

RQD

RQD 100%

SPT…