20c_GR14-15_Geotech_Report.pdf

Geotechnical Report No. 14-15

River ‘S’ Bridge Replacement

FINAL GEOTECHNICAL REPORT

WA RRP RIDG 100(2)

Ridgefield National Wildlife Refuge Ridgefield, Washington

August 2016

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Table of Contents

1. INTRODUCTION

1.1 Project Location

1.2 Scope of Project

2. PHYSICAL SETTING

2.1 Topography, Drainage and Vegetation

2.2 Existing Structure

2.3 Climate

3. SUBSURFACE CONDITIONS

3.1 Site Geology

3.2 Subsurface Investigation

3.3 Subsurface Profile

3.3.1 Fill

3.3.2 Alluvium

3.3.3 Conglomerate

3.4 Groundwater Observations

3.5 Laboratory Materials Testing

4. SEISMIC CONSIDERATIONS

4.1 Site Seismicity

4.2 Liquefaction and Lateral Deformations

5. GROUND IMPROVEMENT RECOMMENDATIONS

5.1 Vibro-Replacement

5.2 Drained Timber Piles

5.3 Ground Improvement Design Methodology

5.3.1 Stone Column Design

5.3.2 Drained Timber Pile Design

6. FOUNDATION ANALYSIS AND RECOMMENDATIONS

6.1 Bridge Foundations Considerations

6.2 Scour

6.3 Bridge Layout

6.4 Design Loads

6.5 Foundation Design Methodology

6.6 Drilled Shaft Recommendations

6.7 Settlement

6.8 Lateral Analysis

7. EXCAVATION AND EMBANKMENTS

8. CONSTRUCTION CONSIDERATIONS

9. LIMITATIONS

10. REFERENCES

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List of Figures Figure 1.1-1 River S Bridge West Abutment Project location maps………………………………..iv Figure 1.2-1 River S Bridge replacement plan.………………………………..……………………………..1 Figure 3.4-1 Vibrating wire piezometer and rain data for boring B15-04vw

Figure 3.4-2 Vibrating wire piezometer and rain data for boring B15-05vw

Figure 3.4-3 Vibrating wire piezometer and rain data for boring B15-10vw

List of Tables Table 2.3-1. Period of Record Monthly Climate Summary for Warren, OR (359051)

Table 3.2-1. Summary of boring locations, elevations, and groundwater observations. .. 6

Table 4.1-1. Site Class Determination Data for Different Areas

Table 4.1-2. Summary of Seismic Parameters (7% PE in 75 years)

Table 4.2-1. Liquefaction Analysis Results

Table 6.8-1. Abutment 1 LPILE soil parameters for lateral analysis

Table 6.8-2. Abutment 2 LPILE soil parameters for lateral analysis

Table 6.8-3. Pier 1 LPILE soil parameters for lateral analysis

Table 6.8-4. Pier 2 LPILE soil parameters for lateral analysis

Table 6.8-5. Pier 3 LPILE soil parameters for lateral analysis

Table 6.8-6. Pier 1 seismic p-modifiers

Table 6.8-7. Pier 2 seismic p-modifiers

Table 6.8-8. Pier 3 seismic p-modifiers Table 6.8-9. Pier 3 residual strengths Table 7.2-1. Abutment 2 settlement estimates

Appendices

APPENDIX A

Site Photographs

APPENDIX B

Site and Exploration Plan

APPENDIX C

Unified Soil Classification System Soil Descriptive Terminology Boring Logs B13-01 to B13-03, B13-04vw, B13-05a, B13-05vw, B13-07 to B13-09, B13-10vw, B13-11, B13-15, B13-16 CPT Logs CPT13-12 to CPT13-14, CPT13-18 to CPT13-21

APPENDIX D

Laboratory Test Reports by GRI Soil Index Test Results by WFL Consolidation Test Results by WFL and WSDOT Direct Shear Test Results by WFL

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Triaxial Compression Test Results by WSDOT

APPENDIX E

Design Axial Geotechnical Resistance Charts Liquefaction Analyses Results Ground Improvement Zone Plan

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Figure 1.1-1. River S Bridge West Abutment Project location maps.

N

River S Bridge

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1. INTRODUCTION

This report provides geotechnical investigation results, analysis, and foundation recommendations for the replacement of the River ‘S’ Bridge. The project stationing referenced in this report is from the 70% Design Set, dated 2 March 2016. Subsurface investigation borings, cone penetration tests, and laboratory tests on intact samples were completed by the Washington State Department of Transportation (WSDOT) on behalf of Western Federal Lands Highway Division (WFLHD). Boring logs, lab test selections, and preliminary analyses were prepared by Geotechnical Resources, Inc. (GRI) on behalf of WFLHD. Laboratory soils index testing and final analyses were performed by WFLHD. This report supersedes all other previous reports for this project, including GRI’s report titled, “Geotechnical Report, River S Bridge Replacement, Ridgefield Wildlife Refuge, Ridgefield, Washington” dated September 16, 2015.

1.1 Project Location

The River S Bridge is located approximately one-half mile south of the City of Ridgefield, Washington. The project location is shown on the maps in Figure 1.1-1 on page iii at the beginning of this report. The bridge crosses Lake River and provides access from the mainland to the peninsular River ‘S’ Unit of the Ridgefield National Wildlife Refuge. The approximate coordinates for the center of the existing bridge are latitude 45.807386° North by 122.740533° West (WGS84).

1.2 Scope of Project

The existing timber trestle bridge was originally built in 1970 and a removable main span composed of steel beams was added in 1980. This single-lane bridge is being replaced due to the poor condition of the existing structure. The replacement bridge will be constructed on a new alignment immediately upstream (south) of the existing bridge within the U.S. Fish and Wildlife management area, as shown in Figure 1.2-1. The existing bridge will be used to carry public traffic during construction, except for periodic closures.

Figure 1.2-1. River S Bridge replacement plan.

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The new bridge superstructure will have a cast-in-place concrete deck supported on prestressed concrete girders. The substructure will consist of drilled shaft foundations, including dual shafts with pile caps at abutments and three single column piers with hammerheads. The bridge will have four spans of 130 feet each for a total length of 520 feet.

Construction of the west abutment will include ground improvement for mitigation of liquefaction and lateral flow. An approximately 30-foot tall MSE retaining wall will be constructed on the eastern approach in order for the new bridge to span over the existing BNSF railroad tracks.

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2. PHYSICAL SETTING

2.1 Topography, Drainage and Vegetation

The Ridgefield Wildlife Refuge Road begins on the east side of Lake River at the intersection with Hillhurst Road at about elevation 265 feet. It follows a natural drainage down to Lake River to about elevation 15 feet. Before the road crosses the river, it crosses a double set of railroad tracks. The Refuge Road crosses Lake River with a timber one lane bridge. The elevation of the roadway is approximately at elevation 38 feet at the top of the existing west abutment and embankment. The roadway slopes down gently to an elevation of approximately 30 feet towards the end of the embankment on the west, with approximately 1V:2H (Vertical:Horizontal) side slopes. The refuge itself is relatively flat in the approximately 10- to 15-foot elevation range, and the peninsula forms the east bank of the Columbia River in this reach. Please note that all elevations in this report are based on the North American Vertical Datum of 1988 (NAVD 88), unless otherwise stated.

The Columbia River and Lake River stages are tidally influenced. Combined with the relatively low elevations of the River S Unit, the area is prone to flooding. We understand that Ordinary High Water (OHW) is at 11.7 feet, and the 100-year flood elevation is at approximately 27 feet. The record high stage at the Vancouver, Washington river gage is reported to be 31 feet in the National Weather Service datum, which correlates to an approximate stage of 36 feet in the NAVD 88 datum (National Weather Service, 2015).

Site photographs are attached in Appendix A. Except for the existing west embankment fill, the refuge area is relatively flat to gently sloping down to Lake River in the proposed improvement area. The ground surface is covered with grass, brush, and several trees on the north side of the embankment. The ground surface is reportedly very soft during the wet winter months and is typically firm during the dry summer months.

The east abutment area is very steep, on the order of 1V:1.2H with an approximately 30-foot wide bench approximately 15 feet below the top of the bank near the proposed Pier 3 location. The slope is heavily vegetated with trees and brush. The close proximity to the railroad right-of-way will also restrict access and staging areas.

2.2 Existing Structure

The existing 331.5-foot long timber trestle bridge was originally built in 1970 with 19 spans varying in length from 8 to 17 feet. A removable 55.5-foot long main span composed of steel beams was added in 1980. The trestle bents and abutments contain timber piles, bracing and caps. The bents and abutments are each supported by 3 timber piles. This single-lane bridge is being replaced due to the poor condition of the existing structure.

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No overhead utilities were observed within the limits of the existing or proposed bridges. We understand that a buried telephone line is within the west embankment and is hung from the bridge. The contractor should verify all utilities prior to construction.

2.3 Climate

The project is located southwest Washington with moderate to heavy rainfall occurring mostly from September through June. Table 1 summarizes average temperatures, precipitation, and snow depth measurements from the Warren, Oregon weather station, located approximately 5 miles west of the site (Western Region Climate Center, 2015).

Table 2.3-1. Period of Record Monthly Climate Summary for Warren, OR (359051)

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec Annual

Average Max.

Temperature

(F)

44.6 50.5 54.7 60.5 68.1 73.2 80.2 79.9 75.6 64.0 52.7 46.2 62.5

Average Min.

Temperature

(F)

32.2 34.1 35.1 38.0 43.7 48.9 52.3 52.4 48.4 42.3 36.6 33.8 41.5

Average Total Precipitation

(in.)

7.87 4.97 4.68 2.60 2.03 1.58 0.50 1.08 1.63 3.73 6.38 7.99 45.04

Average Total SnowFall (in.)

3.8 0.4 0.8 0.0 0.0 0.0 0.0 0.0 0.0 0.0 0.1 1.4 6.6

Average Snow Depth (in.)

1 0 0 0 0 0 0 0 0 0 0 0 0

Notes:

Period of Record: 08/06/1950 to 07/31/1976.

Percent of possible observations for period of record: Max. Temp.: 98.2% Min. Temp.: 98.3% Precipitation: 99.3% Snowfall: 94.8% Snow Depth: 95.4%.

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3. SUBSURFACE CONDITIONS

3.1 Site Geology

The site is within the geologic province known as the Puget-Willamette Lowland which extends from Puget Sound into west-central Oregon between the Coast Range and the Cascade Range. Ridgefield is near the deepest part of the basin which is believed to have been filled with as much as 550 meters (1,800 feet) of sediments carried in from the east by the Columbia River (Evarts, 2004).

The surficial geology of the site is mapped and described on the Geologic Map of the Ridgefield Quadrangle, Clark and Cowlitz Counties, Washington (Evarts, 2004).

Descending from the intersection with Hillhurst Road, the soils along the access road are mapped as cataclysmic-flood deposits (Qfs), conglomerate (QTc), and alluvium (Qa). The cataclysmic flood deposits are described as unconsolidated clay, silt, and fine to medium sand. The conglomerate is described as semi-consolidated pebble and cobble gravel that is poorly sorted to moderately well-sorted and contains minor lenses of cemented sand generally less than 2 meters (6.6 feet) thick. The conglomerate is often locally included as part of the Troutdale Formation. The alluvium is described as unconsolidated sediments underlying the modern floodplains of the Columbia River and Lake River.

The Liquefaction Susceptibility Map of Clark County, Washington (Palmer, 2004a) maps the refuge area west of Lake River as having a moderate to high liquefaction susceptibility. The eastern bank of Lake River is mapped as having very low liquefaction susceptibility. The Site Class Map of Clark County, Washington (Palmer, 2004b) maps the western bank of Lake River as a site class “C” along the river edge transitioning to site class “C to D” and then to a “D” moving west along the map. The eastern bank is mapped as site class “B to C”. Additional discussions on liquefaction and site class are contained in Section 4.

3.2 Subsurface Investigation

A total of 20 explorations were performed for this project, including 13 borings and seven cone penetrometer tests (CPTs). The exploration locations are summarized on Table 3.2-1 and shown on the Site and Exploration Plan in Appendix B. Logs for the borings (B13-01, B13-02, B13-03, B13-04vw, B13-05A, B13-05vw, B13-07, B13-08, B13- 09, B13-10vw, B13-11, B13-15, and B13-16) and graphic summaries of the CPTs (CPT13- 12, CPT13-13, CPT13-14, CPT13-18 through CPT13-21) are contained in Appendix C.

Seven borings were performed in the east bank area, four explorations in the river channel (3 borings and 1 CPT), and nine explorations (3 borings and 6 CPTs) were performed in the west bank area. No CPTs were performed in the east bank area due to the gravelly conditions, which would likely have resulted in refusal in the gravel layers.

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The exploration locations are summarized on Table 3.2-1 and shown on the Site and Exploration Plan in Appendix B.

Table 3.2-1. Summary of boring locations, elevations, and groundwater observations.

Boring No.

Station &

Offset1

Ground Surface

Elevation2 (feet)

Boring Depth (feet)

Water Level Observations2,4

Depth (feet)

Elevation (feet)

B13-01 8+31, 28’ Lt 37.7 160.5 ND 3 ND 3

B13-02 9+85, 13’ Lt 0.2 163.4 0 (in river) River 6

B13-03 11+97, 34’ Lt 42.1 120.5 ND 3 ND 3

B13-04vw 13+65, 36’ Lt 45.3 110.5 19.3-23.6 5 21.7-26.0 5

B13-05vw 8+21, 45’ Rt 15.3 149.5 4.1-14.1 5 1.2-11.3 5

B13-05a 8+35, 53’ Rt 15.3 9.5 2.7 (AD) 12.6 (AD)

B13-07 13+68, 19’ Rt 45.1 102 ND 3 ND 3

B13-08 14+21, 24’ Rt 46.5 49.4 ND 3 ND 3

B13-09 14+17, 21’ Lt 46.5 55.5 ND 3 ND 3

B13-10vw 14+81, 23’ Lt 52.1 53.4 31.5-35.0 5 17.1-20.6 5

B13-11 14+80, 7’ Rt 52.4 49.5 ND 3 ND 3

B13-15 10+04, 28’ Rt 0.5 141.3 0 (in river) River 6

B13-16 11+36, 15’ Rt 7.2 143 ND 3 ND 3

CPT13-12 8+22, 27’ Lt 36.5 112.9 ~30.5 (WD) ~6 (WD)

CPT13-13 8+20, 30’ Rt 16.4 105.3 ~9.2 (WD) ~7.2 (WD)

CPT13-14 10+10, 8’ Lt ~0.2 (est.) 31.0 0 (in river) River 6

CPT13-18 7+54, 33’ Rt 15.6 100.2 ~8 (WD) ~7.6 (WD)

CPT13-19 6+80, 22’ Rt 15.8 29.2 ~7.2 (WD) ~8.6 (WD)

CPT13-20 8+19, 85’ Rt 13.8 25.0 ~6.1 (WD) ~7.7 (WD)

CPT13-21 5+33, 7’ Lt 29.7 132.1 ~22.5 (WD) ~7.2 (WD)

NOTES:

Boring locations and elevations determined by WFLHD Survey.

WD = while drilling, AC = After completion and before casing withdrawal from borehole.

AD = After withdrawal from the borehole. NAVD 88 datum.

Not Determined. Mud rotary drilling methods obscured the water levels.

Water levels at the time of the CPT explorations are estimated from extrapolation of the pore water pressure.

Minimum and maximum recorded range from vibrating wire piezometer over the period from

1/22/2014 to 3/2/2015.

Water elevation equal to Lake River elevation; river level not determined at the time of drilling.

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The borings were completed between January 6 and April 29, 2014. A track-mounted CME 850 drill owned and operated by the Washington State Department of Transportation (WSDOT) was used to advance the borings in the lowland area on the Refuge. A truck-mounted CME 45 drill owned and operated by WSDOT was used to advance the boring on the existing road. Casing advancer methods with a 4-inch diameter casing were used to advance the holes. Where dense gravels and coarser materials were encountered, such as borings B13-07 and B13-16, the drilling method was switched to HQ coring with SPT sampling at the end of the core runs.

Standard Penetration Tests (SPTs) were completed at 2.5-foot to 5-foot intervals in overburden material. The Standard Penetration Test (SPT) consists of using a 140 pound hammer with a 30 inch drop to drive a two inch outside diameter split-barrel sampler 18 inches below the bottom of the bore hole at the sample elevation and recording the number of blows for each six-inch increment of penetration. The last two numbers (for the final 12 inches of penetration) are added together to determine the blows per foot (bpf), which is known as the N-value. Several samples were obtained by driving a modified California sampler with a 2.5-inch O.D., as indicated on the logs. Both of these samplers capture disturbed soil samples. Intact or “undisturbed” soils samples were retrieved with 3-inch diameter thin-wall Shelby tubes per ASTM D 1587.

Borings were logged in the field by a WSDOT Drilling Inspector. Samples were collected and shipped to Western Federal Lands Highway Division (WFLHD) in Vancouver, Washington and WSDOT Materials Laboratory for laboratory testing. The samples and draft boring logs were reviewed by the GRI Geotechnical Engineer assigned to the project. The GRI Geotechnical Engineer reviewing the boring logs also selected samples for laboratory testing.

The CPTs were completed between January 24 and April 30, 2014 by WSDOT using a truck-mounted rig with a Hogentogler 20-ton reaction (push) system. The cone used was a Hogentogler 10 cm2 with four digital channels (tip, friction, pore pressure, and inclination). Shear wave velocity tests were performed at 5-foot intervals. The data acquisition software used was Digital Cone, and the data was post-processed by WSDOT using Coneplot software.

The CPT data is presented in Appendix C as graphic logs provided by WSDOT with tip resistance, sleeve friction, friction ratio, pore pressure, correlated SPT N60 values, and shear wave velocity test results. Also shown are the pore pressure dissipation test results, which were done at select depths in each CPT. We have also provided soil behavior type correlations of the CPT data using the Robertson methods in the CPeT-IT v.1.6.7.42 software by GeoLogismiki.

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3.3 Subsurface Profile

The site is geologically different between the east bank and the west bank. The east bank consists predominantly of loose to very dense conglomerate gravels overlain with varying thicknesses of soft to medium stiff silt/clay alluvium and/or cataclysmic flood deposits. The west bank and river channel consists predominantly of very soft/loose to medium dense silt and sand alluvium. Areas of imported fill and/or regraded native soils cap the areas around the existing west embankment and east approach.

3.3.1 Fill. Material interpreted as fill was encountered in Boring B13-01 to a depth of

18.5 feet below the ground surface (bgs) and within the pre-drill hole for CPT13-12.

These explorations were advanced near the proposed location for Abutment 1. The fill consists of well-graded, subrounded gravel with sand. Field measured SPT N-values were between 11 and 14 blows per foot (bpf). Imported fill or regraded native material is also likely in the upper few feet of borings B13-04vw, B13-07, B13-08, B13-09, B13- 10vw, and B13-11 along the existing east approach.

3.3.2 Alluvium. Alluvial sediments consisting of clay/silt- to sand-sized material were encountered in the borings at the ground surface or below the fill, where fill was encountered. The soil classifications consisted mostly of lean clay, silt, sandy silt, silty sand, and poorly-graded sand. A poorly-graded gravel with sand layer was encountered from 135 to 144 feet bgs in Boring B13-01, and a silty gravel with sand layer was encountered in Boring B13-05vw from 113 to 120 feet bgs.

3.3.3 Conglomerate. Coarse-grained materials consisting of sand- to cobble-sized material in a fine-grained matrix were encountered in the east bank borings below the fill or alluvium, where fill or alluvium was encountered, to the bottom of these borings.

The soil classifications consisted mostly of poorly graded gravel, lean clay with gravel, silty/clayey gravel, silt/sandy silt/silty sand. The dense to very dense gravels at the bottoms of the west bank and channel borings may also be encountering the conglomerate below the alluvium.

The CPT data and soil behavior type correlations indicate that the upper approximately 13 to 32 feet of soil in the CPTs have a more clay like behavior than was logged in the borings. According to Robertson (undated), soft saturated low plastic silts tend to behave more like clays because they have low undrained shear strength and can have a CPT-based Soil Behavior Type (SBT) in Zone 3 (clay). Robertson goes on to state that in general, the normalized chart provides more reliable identification of SBT than the non-normalized charts, although when the in-situ vertical effective stress is between 50 kPa to 150 kPa there is often little difference between normalized and non-normalized SBT.

Therefore, we have also provided normalized charts in Appendix C for comparison. It should be noted here that there are often differences in soil behavior and classification based on the method of determination, such as in-situ tests, laboratory grain size

GR14-15 Page 9 analyses, or determination of the liquid and plastic limits of soil. The boring log soil classifications are based on the soil identification of the person logging the boring and refined with the results from the laboratory index tests. The CPT logs and interpreted information are provided as supplemental information for other geotechnical design purposes, such as ground improvement and pile driving.

3.4 Groundwater Observations

Water level observations made in the subsurface investigation borings are summarized in Table 3.2-1. Due to the shallow groundwater and loose granular deposits, the borings used casing advancer and coring drilling methods to reduce heave and caving within the boring. The addition of drilling fluids in both these methods prevents direct water level measurements while drilling. Only Boring B15-05a, which was performed only to a depth of 9.5 feet bgs for shear vane testing, was drilled without mud making direct water level measures possible after removing the casing.

Groundwater levels were also estimated from the CPT pore pressure measurement and dissipation data. These groundwater levels should be considered approximate and are reported on Table 3.2-1.

To monitor the groundwater level over seasonal fluctuations, a vibrating wire (VW) piezometer was installed in borings B13-04vw, B13-05vw, and B13-10vw. The pressure and temperature was recorded from January 2014 to March 2015 and converted to groundwater elevations. The charts below plot the groundwater elevation over the recorded time period. Precipitation data is also plotted for reference.

The groundwater table is expected to be influenced with the adjacent Lake River water surface elevation, which does at times inundate the area. Subsurface water is likely to be encountered during drilled shaft construction at or near the elevation of the river.

Water elevations are likely to fluctuate with fluctuations of the Lake River, particularly during spring thaw and following periods of heavy precipitation. Lake River and the Columbia River are also tidally influenced.

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Figure 3.4-1 Vibrating wire piezometer and rain data for boring B15-04vw

Figure 3.4-2 Vibrating wire piezometer and rain data for boring B15-05vw

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Figure 3.4-3 Vibrating wire piezometer and rain data for boring B15-10vw

3.5 Laboratory Materials Testing

Selected split-barrel (SPT) soil samples and intact Shelby tube samples were submitted to the WFLHD Materials Laboratory for routine index testing and direct shear testing.

Several intact Shelby tube samples were submitted to the WSDOT Materials Laboratory for one-dimensional consolidation tests and one sample was submitted for a consolidated undrained triaxial compression test. Laboratory test reports are contained in Appendix D.

The following is a list of tests that were completed on the selected soil samples and the standards that were followed:

Particle Size Analysis – AASHTO T 88

Liquid Limit, Plastic Limit, and Plasticity Index – AASHTO T 89/90

Natural moisture – AASHTO T 265

Apparent Specific Gravity – AASHTO T 100

One-Dimensional Consolidation – AASHTO T 216

Direct Shear – AASHTO T 236

Consolidated Undrained Triaxial Compression – AASHTO T 297

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The results of the soil index tests were used to classify the samples according the Unified Soil Classification System (USCS). The USCS classification of individual laboratory tested soil samples is reported on the boring logs.

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4. SEISMIC CONSIDERATIONS

4.1 Site Seismicity

The procedures outlined in Section 3.10 of the 7th Edition (2014) of the AASHTO LRFD Bridge Design Specifications (BDS) were followed to characterize the seismic hazards.

These procedures use a design earthquake with a 7 percent probability of exceedance in 75 years, which equates to a return period of approximately 1,000 years.

Information from the USGS Earthquake Hazards Program website was used to locate nearby faults. According to USGS information, the closest active faults are the Portland Hills fault approximately 11 miles southwest of the project and the Lacamas Lake fault approximately 14 miles southeast of the project. Since no active faults are located within 6 miles of the bridge, additional studies do not need to be performed to quantify near fault effects on ground motions per AASHTO.

As discussed previously, the geology of the west bank and east bank are appreciably different. The west side subsurface soils consist predominantly of silt to sand sized alluvial sediments, where the east side consists of some fine-grained alluvial/colluvial sediments underlain by relatively shallow gravelly conglomerate. As discussed in the Geology section, the west bank is mapped as site class “C”, where the east bank is mapped as site class “B to C”. Although site class maps are used for planning purposes only, it is important to note the differences in site classes based on the geologic formations.

Based on the AASHTO procedures, the shear wave velocity, standard penetration resistance, or undrained shear strength in the upper 100 feet of the soil profile can be used to determine the site class. Seismic site classes range from A to F and are arranged in order of decreasing density of subsurface materials and increasing susceptibility to structural damage as a result of an earthquake. The average shear-wave velocity in the upper 100 feet collected from the CPTs on the west side of the river is about 598 feet per second or 182 meters per second; no CPTs were performed on the east side due to the likelihood of refusal in the gravelly and denser soil conditions there. The cutoff between site class D and E is listed in AASHTO as 600 ft/s (183 m/s), but is listed as 180 m/s in NEHRP (2003) and other site class publications using metric units. This rounding to two significant figures shows the west side shear wave velocity values are on the cutoff line. If shear wave velocity data were obtained on the east bank, the denser and gravelly soil conditions would likely be within the site class D range.

The average N60 value for the upper 100 feet of borings on both sides of the river is 13, but if broken down per area as shown in the table below, each area can be classified as a different site class. These blowcounts are close to the cutoff of 15 blows per foot between site classes D and E. It should be noted that other publications listing site

GR14-15 Page 14 classification ranges, such as International Building Code and NEHRP, state that the blowcount value, N, should be determined on the values measured in the field without correction. The AASHTO Bridge Design Specifications (BDS) do not state what N value to use. However, based on internal discussions at FHWA, the N60 value (field N corrected for hammer efficiency), but not the (N1)60 value (N60 corrected for overburden pressure and other factors), should be used for site classification purposes and may be updated in future editions of the AASHTO BDS.

Table 4.1-1. Site Class Determination Data for Different Areas

Area Avg. N60 or Vs Upper 100 ft Site Class

West Bank Borings 8 blows/ft E

West Bank CPTs 598 ft/s (182 m/s) D/E

In River 10 blows/ft E

East Bank Borings 16 blows/ft D

Since the worst site conditions from a seismic perspective were at the western side where shear wave velocities were measured at the lower end of the site class D range, then it could be argued that the overall site class across the bridge should be within the site class D range with soil conditions improving towards the east. Furthermore, shear wave velocity testing is the preferred method (AASHTO BDS Table C3.10.3.1-1 Method A) over determination by blowcount (Method B) or undrained shear strength (Method C). Others could conservatively argue that the site class should be based on the lower site class E, since there are no direct shear wave velocity measurements elsewhere and therefore, the SPT blowcounts should be relied upon. It is the opinion of the WFLHD that, based on all the available data, the site class for this bridge design should be considered a borderline “D/E” and that site acceleration coefficients could be more accurately determined by linearly interpolating halfway between the two site classes in Tables 3.10.3.2-2 and 3.10.3.2-3 of the AASHTO BDS. The AASHTO BDS does not specifically state to interpolate between classes; however, the BDS does state to linearly interpolate between PGA values in the tables. Even though soil behavior is generally non-linear as discussed in the BDS, interpolation between intermediate table values is generally recommended throughout the above-referenced NEHRP document.

The USGS interactive deaggregation website (2008) was used to determine the peak bedrock acceleration and spectral accelerations at the project location. The spectral accelerations were then factored to Site Class D and Site Class E soil conditions using the site coefficients in the AASHTO BDS. The resulting values were then averaged between the two site classes. The results of this analysis are presented in Table 4.1-2.

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Table 4.1-2. Summary of Seismic Parameters (7% PE in 75 years)

Recommended Site Class D/E

Horizontal Peak Ground Acceleration on Site Class B Rock (PGA)

0.260 g

Horizontal Response Spectral Acceleration at a

Period of 0.2 sec on Site Class B Rock (Ss)

0.613 g

Horizontal Response Spectral Acceleration at a

Period of 1.0 sec on Site Class B Rock (S1)

0.221 g

Average D/E PGA site coefficient, (FPGA) 1.34

Average D/E Short Period Site Coefficient, (Fa) 1.39

Average D/E Long Period Site Coefficient, (Fv) 2.54

Factored Horizontal Response Spectral Acceleration at a

Period of 0.2 sec for Project Site, Site Class D (SDS)

0.852 g

Factored Horizontal Response Spectral Acceleration at a

Period of 1.0 sec for Project Site, Site Class D (SD1)

0.561 g

AASHTO Seismic Zone 4

4.2 Liquefaction and Lateral Deformations

Soil liquefaction is a phenomenon in which cohesionless soil deposits below the groundwater table lose a substantial amount of strength due to pore pressure generation resulting from earthquake strong ground shaking (Kavazanjian, 2011). A liquefaction analysis was performed using the guidelines provided in the AASHTO BDS and the 2013 Washington State Department of Transportation (WSDOT) Geotechnical Design Manual (GDM). Notably, the procedures discussed in Appendix 6-A.2 of the WSDOT GDM were used to perform the liquefaction using the WSDOT software WSLiq developed by Professor Steve Kramer of the University of Washington. This software performs the recommended deaggregation analyses, considers multiple earthquake sources, and incorporates several different methods of analyses using the WSDOT recommended weighting. The results of the liquefaction analyses for the West Abutment (Abutment 1) are summarized in Table 4.2-1 below.

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Table 4.2-1. Liquefaction Analysis Results

Exploration No.

Depths of Layers Predicted to Liquefy1 (ft)

Estimated Seismic Settlement (ft)

Estimated Lateral Spreading2 (ft)

B13-01 28 - 80 1.1 5.2

B13-02 0 - 50 1.9 NE

B13-03 28½ - 33½, 43 - 58, 63 - 68 0.4 1.5

B13-04vw 37 - 42, 52 - 57 0.2 NE

B13-05vw 10 - 62½, 72½ - 80 1.3 7.4

B13-07 23 - 30 0.2 1.3

B13-15 0 - 24, 28 - 63, 68 - 80 1.4 NE

B13-16 0-17, 22-27, 33-43, 46-68½ 1.1 2.6

CPT13-12 33 - 80 1.3 17

CPT13-13 15 - 80 1.7 20

Note:

Maximum considered liquefaction depth was limited to 80 feet per WSDOT GDM.

Not expected.

Since the CPT data could not be directly used in the WSLiq software, these were not analyzed using the multiple source methodology developed by Kramer. The CPT results used the weighted Mw of 6.99 and a weighted As of 0.27 g in the Robertson modified NCEER (National Center for Earthquake Engineering Research) analysis mode of the software CLiq by Geologismiki.

The results of the analysis indicate that material with the potential to liquefy under the considered seismic events extends deeply and will result in significant vertical and horizontal displacements, including lateral flow of the west bank beyond the estimated lateral spreading displacements. To mitigate these displacements on the west bank, we recommend a ground improvement program to strengthen and/or reinforce the soils in a zone around the Abutment 1 drilled shaft foundations to reduce the liquefaction potential and resulting lateral and vertical displacements.

The east bank lateral spread displacements were modeled using procedures described by Ashford, et al. (2013). The residual strengths of the liquefiable layers were determined from the WSLiq analysis and slope stability of the section through Pier 3 modeled in SLIDE v6.0 software by Rocscience. The factor of safety was estimated to be less than 1.0, which indicates a flow type failure. The lateral flow conditions were then modeled using LPile v6.0 software by Ensoft, Inc. WFL provided the LPile input file to HDR for determination of adequate structural capacity. For structural analysis, Ashford states that the likelihood of the peak shear force and peak displacement demand occurring at the same time is low, therefore, only 50 percent of the inertial load from the superstructure and column should be combined with the kinematic loading. Based on the lateral parameters provided by WFL, HDR determined that the drilled shaft for Pier 3 has adequate capacity under the extreme earthquake event from a collapse prevention standpoint. Therefore, ground improvement is not required at the east bank

GR14-15 Page 17 for support of the bridge structure. Please note that ground deformations will still likely occur, which may result in structural deformations that require repair or closing/rebuilding of parts or all of the bridge during an extreme earthquake event. It is our understanding that this low traffic volume bridge is only to be designed for collapse prevention and not serviceability immediately following a design earthquake, consistent with the seismic design philosophy of the AASHTO LRFD Bridge Design Specifications.

Liquefaction potential and associated settlements for the Abutment 2 (east) is estimated to be low and within the structural tolerance of the abutment shafts. No mitigation is recommended for Abutment 2. Pier 1 and Pier 2 will likely experience liquefaction, but low to no lateral spreading since they are in the relatively flat river channel bottom. Downdrag loads have been considered in the axial capacities of Pier 1 and Pier 2 drilled shafts.

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5. GROUND IMPROVEMENT RECOMMENDATIONS

Ground improvement is a generic term for improving the density and/or strength of a soil to reduce static settlements, seismic settlements, lateral displacements, and other deformations. Ground improvement methods may include deep dynamic compaction, rapid impact compaction, dry and wet soil mixing, rigid inclusions, vibro-densification, vibro-replacement, and grout/injection methods, among many others. We recommend two methods be considered for design of this project: 1) vibro-replacement methods;

and 2) rigid inclusions using drained timber piles.

5.1 Vibro-Replacement

Vibro-replacement methods are also known as stone columns or aggregate piers, as well as other proprietary names. For the purposes of this report, vibro-replacement is defined as a method of replacing in-situ soils with angular gravel and compacting the gravel into a dense column or pier. Installation methods typically involve the use of a horizontally vibrating or oscillating probe called a vibroflot, or a vertical ramming mandrel with a beveled compacting tip. Gravel is typically fed by gravity, compressed air, or water to the bottom of the cavity through a tremie pipe attached to the probe or mandrel. As the probe is lifted, gravel fills the cavity, and then is compacted as the probe is lowered and vibrated or rammed. Drill and fill methods are also available where a hole is drilled out and then filled with the stone and compacted in lifts. However, the use of drill and fill methods is typically not applied where shallow groundwater conditions, such as at the West Abutment site, can cause caving and heaving.

For liquefaction mitigation, the stone column itself is not a structural element, but if spaced closely enough in an area, the stone columns increase the strength of the composite soil matrix thereby improving the ground. The native matrix soil is also improved to varying degrees by displacement from the probe or densification from the vibratory actions or ramming. Typically, sandy soils with less than 15 percent passing the No. 200 sieve can be densified by the vibratory action, whereas soils with more than 25 percent passing the No. 200 sieve typically do not see much densification from vibratory action. The bulging of the stone column into the softer matrix soils during compaction also provides lateral stress increases into the soils and a corresponding increase in soil strength.

5.2 Drained Timber Piles

Drained timber piles consist of timber piles used in conjunction with prefabricated vertical drains (PVDs). The timber piling is typically spaced 3 pile diameters on center to create a soil-pile composite with increased shear strength and stiffness. The PVDs, also known as wick drains, are thin composite geotextile drains consisting of a channelized plastic strip sandwiched between two geotextile filter fabrics. The PVDs are typically 4

GR14-15 Page 19 inches wide by 1/8 to 1/4 inch thick. By installing PVDs at relatively close horizontal spacing, such as 3 to 7 feet on center, the PVDs can provide a shortened drainage path in slowly consolidating soils to speed the rate of settlement. PVDs have also been used in ground improvement installations to improve the ability of finer grained soils to densify from the displacement and compactive efforts by allowing the excess pore water pressures to dissipate more quickly.

With drained timber piles, the PVDs are typically attached directly onto the sides of the timber piles to allow the excess pore pressures generated from driving displacement to dissipate quickly. By allowing the pore water pressures to dissipate during driving, the volume of the soils is reduced resulting in greater densification and strength increases.

PVDs can also be installed using typical wick drain installation mandrels; this separate installation would likely result in added costs, although it may benefit drivability and ease installation of the piles.

5.3 Ground Improvement Design Methodology

Ground improvement as specified for this project will consist of installing either stone columns (including aggregate piers or other named methods meeting the specifications) or drained timber piles to depths of approximately 60 feet below elevation 19 feet (tip elevation of -41 feet), which is the planned work platform elevation in the area of the new west approach embankment.

Some liquefaction-induced settlement up to approximately 9 inches could still occur below the ground improvement zone at the west abutment area; however, it is not considered feasible nor necessary to reduce the liquefaction induced displacements to zero.

5.3.1 Stone Column Design

The primary intent of the stone column ground improvement is reinforcement of the composite soil structure with the addition of the gravel matrices, particularly in regard to reducing lateral deformations and bank instability due to residual soil strengths after an earthquake. Our slope stability analyses using residual soil strengths outside the ground improvement zone show factor of safety greater than unity from slope failure up to the abutment structure for an area replacement ratio of 0.2. Some slope failure between the abutment and river are assumed to occur, particularly outside or through the edges of the ground improved zone. Secondary benefits of ground improvement are densification of the in-situ sandy soils and improved drainage paths to reduce liquefaction potential and provide lateral resistance equal to or greater than static parameters. Since a significant thickness of the soils is fine-grained at or just below the river bank where slope instability is critical, we cannot depend on densification to provide the mitigation for lateral stability. Therefore, the specifications do not require

GR14-15 Page 20 densification to be a performance specification. However, we do require pre- and post-ground improvement CPT testing to document the constructed improvement.

5.3.2 Drained Timber Pile Design

The primary intent of drained timber piles is to densify the soils through volume displacement and the large driving forces with pore pressure dissipation through the PVDs. Although design methodologies have not been refined for this ground improvement method yet, recent yet to be published research shows that sufficient ground improvement can be obtained at a horizontal spacing of 3D, where D is the pile butt diameter, on a triangular spacing pattern. An added benefit is the shear reinforcement provided by the timber pile structure as a rigid inclusion. These timber elements would provide additional shear resistance within the ground improvement zone.

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6. FOUNDATION ANALYSIS AND RECOMMENDATIONS

The following analysis and recommendations reference the bridge design by HDR, Inc., Portland, Oregon. It should be restated here that the ground improvement design for liquefaction mitigation only partially mitigates the estimated liquefaction hazard around the Abutment 1 area since it is not feasible to fully mitigate the hazard. Some settlement on the order of 9 inches could occur below the ground improvement zone following the design earthquake event. The ground improvement design is primarily intended to reduce lateral flow instability in the river bank and channel elevations, as well as provide static load capacities for the drilled shafts. Therefore, it is our intention that the bridge designer details the structure with ductile connections capable of tolerating the expected displacements for the no collapse criteria. As discussed in Section 4.2, our analyses show that the drilled shaft at Pier 3 should be capable of tolerating the estimated lateral displacements without ground improvement, and little to no displacements are anticipated at the eastern Abutment 2.

6.1 Bridge Foundations Considerations

Site conditions, depth of liquefaction, scour, site seismicity, design loading, cost, and constructability were the major factors considered when evaluating the foundation alternatives. Shallow foundations were considered and dismissed as a foundation option due to the relatively high seismic loads that are anticipated, the potential for lateral flow, and the relatively large excavations and associated impacts that would be required adjacent to the existing bridge.

Driven piling was considered but later eliminated from consideration for the following reasons:

Large axial and lateral loading would require an excessively large pile group

Pile driving in the salmonid bearing river would result in construction restrictions and added costs, such as bubble curtains, shorter work windows, etc.

Pile tip elevations can be difficult to predict in these types of soil conditions on the west side and in the river, and pile driving on the east side can be obstructed by the gravel layers over weaker substrata.

Drilled shafts will be able to support relatively large axial and lateral loads. Drilled shafts also require a minimal construction footprint in the river and can sometimes be constructed without a cofferdam. Use of drilled shafts has much less uncertainty in estimated tip elevations than driven piling. Only one foundation type at all bent locations is preferred for bridge design.

Based on these considerations drilled shaft foundations are recommended for support of the new bridge.

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6.2 Scour

Megan Frye, WFLHD Hydraulics Engineer, provided the following recommendation for scour in the River S Bridge TS&L Hydraulics Report, dated June 17, 2014:

The primary mode of scour expected at the proposed bridge is pier scour for the 100-year and 500-year flood events at 3.61 feet and 3.74 feet respectively.

Although no scour at the abutments was determined in the calculations, the flow conditions at the 100-year and 500-year flood are difficult to represent accurately in the calculations without a more comprehensive two-dimensional flow analysis as the flow will be divided in both the channel and on the west floodplain. Given the relatively flat gradient and slow velocities, the expected conditions at the west abutment can be approximated without an in depth modeling effort.

Based on the scour estimates above and consideration of disturbance of the soils in the upper 5 feet due to construction, skin friction and lateral capacity was ignored in the upper 5 feet.

6.3 Bridge Layout

The new bridge structure will have the following characteristics:

Four spans of approximately 130 feet each

Offset up to approximately 50 feet south of existing bridge

Total length of 520 feet

Total width of 24 feet

Approximate skew angle of 0 degrees

At each abutment the superstructure will be supported on two 5-foot diameter drilled shafts and pile caps. At the pier locations the superstructure will be supported by a single 10-foot diameter drilled shaft and pier column with a hammerhead.

6.4 Design Loads

HDR provided the following factored axial compressive loads for the abutments and piers:

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Table 6.4-1 Vertical Load Demand at Top of Shaft

Service I (kips)

Strength I (kips)

Extreme Event I (kips)

Abutment 1 479 654 700

Abutment 2 464 681 691

Pier 1 1803 2473 1770

Pier 2 1817 2488 1786

Pier 3 1861 2546 1842

For Abutment 1, downdrag loads could also be imposed from static settlement from embankment construction after installation of the drilled shafts. We estimate the magnitude of the downdrag may be up to approximately 520 kips (including a load factor of 1.25 per AASHTO BDS) for the 5-foot diameter shaft due to settlement of the very soft silt material below the new embankment between elevations 0 to 12 feet. The shaft tip elevation of -50 feet includes enough geotechnical axial capacity to accommodate downdrag forces in both Service and Strength Limit States.

Downdrag on Abutment 1 and Piers 1 to 3 may also occur due to seismic settlement from liquefaction. Downdrag loads develop when soil settles and pulls down on a drilled shaft that is restrained from movement at the tip. However, since ground improvement will be installed along most of the length of the Abutment 1 drilled shafts, liquefaction settlements are expected to occur near and below the shaft tips and the entire ground improvement zone should settle more or less as a monolithic unit. The tip of the drilled shafts will be founded in a layer of medium dense poorly graded sand. Drilled shafts bearing in granular soils must displace downward more than 10 percent of the base diameter to mobilize the full base resistance. Therefore, downdrag loads should not develop in the shaft since the tip of the shaft will not likely be restrained by a hard bearing layer. Per AASHTO BDS section 10.8.1.6.2, if the magnitude of the ground surface liquefaction settlement is less than the failure criterion for shafts with tip bearing in soil, then the liquefaction downdrag, if any, can be ignored for strength and extreme limit states. This criterion applies at Abutment 1, and it is our understanding that the bridge structure can tolerate the expected liquefaction settlements discussed previously in this report.

The bottoms of the drilled shafts for Piers 1 to 3 bear in soils deep or dense enough to induce downdrag loads. Therefore, downdrag loads have been considered in the reduced capacities for the extreme liquefaction event and are shown on the Design Axial Geotechnical Resistance Charts in Appendix E. Since these drilled shafts will likely bear in non-liquefiable or dense layers, excessive settlements of the structure are not anticipated, although the increased demands on the pile integrity (i.e., potential for buckling or shear) should be checked by the structural engineer.

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6.5 Foundation Design Methodology

Drilled shaft foundations were designed in accordance with the 7th Edition of the AASHTO LRFD Bridge Design Specifications (2014). Drilled shafts were designed using a combination of skin resistance and tip resistance based on a 5-foot abutment shaft diameter and the 10-foot pier shaft diameter. Side resistance should be reduced or neglected within the zone of permanent casing. Both axial compressive and uplift resistance were evaluated at the abutments. Charts of factored axial resistance versus depth are provided in Appendix E.

6.6 Drilled Shaft Recommendations

Recommendations for drilled shafts are summarized in Table 6.6-1. In addition to the design loads, subsurface conditions and settlement criteria were considered when establishing the recommended drilled shaft tip elevations. Where the resistance factors are to be applied to a single shaft supporting a bridge pier, the resistance factor values should be reduced by 20 percent.

Table 6.6-1. Drilled shaft recommendations.

Abutment 1 Abutment 2 Pier 1 Pier 2 Pier 3

Station at Centerline of Bent

8+20 13+40 9+50 10+80 12+10

Top of Drilled Shaft Elevation

24 ft 41 ft 12 ft 12 ft 30 ft

Number of Drilled Shafts per Bent

2 2 1 1 1

Diameter 5 ft 5 ft 10 ft 10 ft 10 ft

Bottom of Shaft Elevation -50 ft -20 ft -120 ft -105 ft -65 ft

Total Length of Drilled Shaft

74 ft 61 ft 132 ft 117 ft 95 ft

Geotechnical Resistance Factors -

Side resistance / tip resistance

0.55/0.50 0.55/0.50 0.55/0.50 0.55/0.50 0.55/0.50

Nominal Axial Compressive Resistance

2,200 kips…