Dollar_Creek_Bike_Path_Geotechnical_Memo_09-10-14.pdf

To:

From: Khamis Y. Haramy, Lead Geotechnical Engineer, CFLHD, Lakewood CO

Through: Matt Ambroziak, Project Manager, CFLHD, Lakewood CO

Subject: Geotechnical Design Recommendations for the Dollar Creek Shared-Use Trail

Project, CA FLAP 10089(1), Placer County, and Lake Tahoe District, CA

BACKGROUND

The Central Federal Lands Highway Division (CFL) of the Federal Highway Administration (FHWA), in cooperation with Placer County Department of Public Works and Tahoe Regional Planning Agency (TRPA), is proposing to construct 2.2 miles of new shared-use path along the north shore of Lake Tahoe near State Route (SR) 28 between Dollar Hill and Tahoe Vista. The project will include a shared-use trail, a bridge over Dollar Creek, a marked pedestrian crossing at SR 28 and Country Club Drive, a parking area off SR 28, and a neighborhood connector at County Club Drive. The Dollar Creek shared-use trail begins at the Dollar Hill area northeast and ends at Fulton Crescent Drive. The bicyclists are not expected to reach speeds greater than 20 mph. The trail will be a paved ten-foot wide pedestrian path with a two-foot clear zone on each side.

The objective of this memorandum is to provide recommendations regarding structure foundations for the proposed bridge based on field observations and available soil data published in the environmental documents and the scoping report(s). Due to the low risk and budget constraint, no subsurface investigation was conducted at this site by CFL. Available data interpretation and engineering analysis was conducted using conservative assumptions by geotechnical personnel from CFL.

REGIONAL GEOLOGY

Lake Tahoe is located on the north portion of the Sierra Nevada Geologic Province, specifically the Tahoe-Truckee Sub-section. The area is also at the margin of two major geologic regions; the Sierra Nevada, and the Basin and Range Geomorphic Regions. The two regions have distinct geologic, geomorphic, hydrologic, and tectonic characteristics. The regional geologic setting of the project area is characteristic of the Sierra Nevada region with steep topography developed on primarily granitic bedrock of the Sierra Batholith. The Batholith represents a series of large igneous intrusions emplaced during the Paleozoic Era (575 to 270 million years ago). The Sierra Nevada mountain region has been uplifted as a tectonic block. Portions of the Basin are mantled by glacial till (generally undifferentiated mixtures of gravel and boulders in a matrix of finer-grained sediments) deposited by a succession of glaciations during the Quaternary period (last 2 million years). The larger stream valleys are partially filled with glacial outwash sediments (silt, sand, and some gravel) and more recent lacustrine and alluvial deposits.

Lake levels have fluctuated within Lake Tahoe Basin in response to climatic changes and outlet restrictions from the lake into the Truckee River. Geologic evidence indicates that historically lake levels may have been significantly higher than present levels. Evidence of higher lake levels is reflected in the mapping of Lake Terrace deposits above the current water line on the southern margin of the lake. Recent

Technical Memorandum

Date: August, 10th, 2014 investigations near the southern margin suggest that a relatively high stand of the lake resulted in deposition of lake sediments within the Upper Truckee Marsh during the period 14,000 to 25,000 years ago. The tectonic conditions and geologic structure of the Lake Tahoe basin are characteristic of the Basin and Range region. The Lake Tahoe basin is a fault-bounded valley formed by the extensional tectonic regime which defines the Basin and Range. The valley is the western-most basin of this portion of the Basin and Range and is bounded by the Sierra Nevada to the west and the Carson Range to the east. More specifically, the basin is a half graben, formed between the West Tahoe Fault, located on the western margin of the lake, and the North Tahoe-Incline Village Faults at the north end of the lake.

LOCAL GEOLOGY

Lake Tahoe and the Tahoe-Truckee Sub-section province is a 400-mile tilted fault-block created by tectonic and localized stresses in the earth crust with the eastern face being more rugged and steep compared to the gentle western slope that plummets beneath the adjacent San Joaquin Valley. The Tahoe- Truckee Sub-section contains moderately steep plateaus and steep mountains. Streams form canyons with steep side-slopes. Elevation ranges from 5,000 feet along the Truckee River to 9143 feet at Mount Lola.

The Lake Tahoe Basin was formed by geologic block (normal) faulting that occurred about 3 million years ago. A geologic block fault is a fracture in the Earth's crust causing blocks of land to move up or down. Uplifted blocks created the Carson Range on the east and the Sierra Nevada on the west. Down-dropped blocks created the Lake Tahoe Basin in between. Some of the highest peaks of the Lake Tahoe Basin that formed during this process were Freel Peak at 10,891 feet, Monument Peak at 10,067 feet Pyramid Peak at 9,983 feet and Mt. Tallac at 9,735 feet

Snow, rain, and streams filled the southern and lowest part of the basin, forming the ancestral Lake Tahoe. Modern Lake Tahoe was shaped and landscaped by the scouring glaciers during the Ice Age (the Great Ice Age began a million or more years ago). Many streams flow into Lake Tahoe, but the lake is drained only by the Truckee River, which flows northeast through Reno, NV and into Pyramid Lake in Nevada. Lake Tahoe is the second deepest lake in the United States and the tenth deepest in the world, with a maximum depth measured at 1,645 feet and an average depth of 1,000 feet

GEOLOGICAL HAZARDS

Potential geologic hazards for the project area include proximity to potentially active faults and potential liquefaction resulting from loose sandy soil conditions. The most significant potential geologic hazards associated with the project area are caused by earthquakes and their associated effects. Earthquakes present direct (primary) and indirect (secondary) hazards; both of which can occur locally or at locations distant from the earthquake source. Direct, local earthquake hazards include damage caused by fault displacements either by ground surface rupture or gradual fault creep. The damage caused by ground shaking is also a direct effect; however, shaking can also occur locally or at remote locations. Indirect hazards presented by earthquakes include liquefaction and earthquake-induced landslides, both of which are triggered by ground shaking. Small segments of the project area that are located on or near steep terrain could be subject to earthquake induced slope hazards. Most slopes within the project area are sloping at flatter than 30 percent and should not be affected. Roads, structures, pipelines, utilities lines, dams and embankments in the project area vicinity may also be subject to this hazard. The analysis of these hazards is based on an understanding of the potential for these events to occur in the project area.

SEISMIC DESIGN PARAMETERS

Based on a review of the Preliminary Map of Pleistocene to Holocene Faults in the Lake Tahoe Basin, the project area intersects the Dollar Point Fault, an active fault. As a result, the potential exists for displacement caused by Fault Rupture or Creep along some sections of the trail and some slopes in the northern segment of the project area. The project area is mapped as having a probable maximum earthquake magnitude of 7.0 with moderate severity of ground shaking during an earthquake. The project area that crosses Dollar Creek is anticipated to experience the greater severity of ground shaking due to soil conditions.

The American Association of State Highway and Transportation Officials (AASHTO) criteria for seismic design allows for some structural damage to occur when the structure is subjected to 1000-year strong ground motion. For Extreme Event I Limit State Design an earthquake load and resistance factor of 1.00 should be used during the LRFD design of structures.

Recommended seismic response parameters for the Dollar Creek Project design are based on the AASHTO LRFD Bridge Design Specifications, 5th Edition, 2010 and represent horizontal peak ground acceleration with 7 percent probability of exceedance in 75 years (approximate 1000-year return period).

The 1000-year return period data for the project site located at 39.18752 degrees latitude and -120.10253 degrees longitude, was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short- and long- period ground acceleration (Ss and S1 respectively) for a certain soil profile at the bridge site.

The soils near the Bridge foundations mainly consist of the Tahoma and Jorge series overlying andesitic bedrock which predominantly underlies the northwestern flank of the Carson Range. These two soil series are generally sandy soils with cobbles and boulders of varying sizes that occur on gently sloping to steep slopes. The overburden soils are formed from residuum derived from the weathering of volcanic rock (andesite). Large boulders are present on the surface and should be expected during construction.

Subsurface profiles similar to the ones encountered at the project site with soil layers overlaying bedrock that significantly differ in stiffness and density, will have amplified ground motion and resonant period governed by the layer thickness and shear wave velocity of the materials. The damage potential of strong ground motions with respect to a structure is typically affected by the period of the earthquake motion and the resonant period of both soils the soil and the structure. The shear wave velocities in the top 100 feet (Vs100) of the soil/rock were not measured but based on available subsurface data from the average-time-weighted shear velocity for the top 100 feet (Vs100) of the subsurface materials can be conservatively estimated between 1,200 and 2,000 feet per second therefore the site is classified as “Class C” according to the site class definitions specified in Table 3.10.3.1-1 of AASHTO.

A seismic hazard analysis to establish ground motions for seismic design was conducted. The recommended spectral acceleration coefficient values for probabilistic design with a return period of 1,000 years were calculated using the computer program provided with the AASHTO LRFD Bridge Design Manual developed by the USGS (2008) entitled “Seismic Design parameters,” version 2.10 and are summarized in Table 1.

TABLE 1:- Summary of Seismic Parameters for Earthquake with 1000-Year Return Period.

Horizontal Peak Ground Acceleration, (PGA) 0.383g Horizontal Response Spectral Acceleration at Period of 0.2 sec, (Ss) 0.937g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (S1) 0.332g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.02 Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.03 Site Factor at Long-Period Range of Acceleration Spectrum, (Fv) 1.47

Factored Horizontal Peak Ground Acceleration, (As) 0.394g Factored Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDs) 0.965g Factored Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.487g

Seismic Zone Zone 3

The five percent damped design response spectrum for a total period of 2.00 seconds was calculated using spectral acceleration coefficients listed in Table 2.

Based on an acceleration coefficient (SD1, the seismic response motion acceleration at 1.00 second) value of 0.487g calculated as Fv.S1, the bridge is assigned a seismic performance Zone 3 in accordance with Table 3.10.6-1 in AASHTO (0.50 ≤ Fv.S1). Seismic zones reflect the variation in seismic risk in different parts of the country and should be used for permitting different requirements for design. Seismic design forces in the lateral direction for structures shall be determined by dividing forces obtained from Article

3.10.8 by the appropriate response modification factor, R, specified in Table 3.10.7.1-1 in AASHTO.

TABLE 2:- Summary of Design Response Parameters for Site Class C.

TIME, sec Sa, g Sd, in REMARKS

0.000 0.390 0.000 T = 0.0, Sa = As

0.119 0.965 0.096

0.200 0.965 0.377 T = 0.2, Sa = SDs

0.594 0.965 2.405

0.600 0.812 2.857

0.800 0.609 3.809

1.000 0.487 4.762 T = 1.0, Sa = SD1

1.200 0.406 5.714

1.400 0.348 6.666

1.600 0.305 7.619

1.800 0.271 8.571

2.000 0.244 9.523

The magnitude associated with the design ground motion is required for site response or time domain structural analysis. The 2002 USGS Interactive Deaggregation for the Conterminous United States was used to establish a design magnitude of 7.0 earthquakes based on the peak ground acceleration. For consistency with AASHTO, the deaggregation was conducted for a ground motion with a 5 percent probability of occurrence in the 50 year period, a ground motion with a return period of approximately 975 years. This return period is consistent with the 1000 year return period for the AASHTO seismic hazard map.

SUBSURFACE INVESTIGATIONS

One five foot-deep test sample (TP-4 and TP-5) was taken at each abutment location. The encounter subsurface materials were relatively identical from both test sites. The soils were 0.0 to 0.3 foot of topsoil,

0.3 to 3.0 feet sandy silt with gravel, and 3.0 to 5.0 feet of hard sandy gravel with cobbles up to 8 inches in diameter. The materials encountered at the bottom of the test sites was tested in the laboratory for gradation/classification and classified as A-1-b (0) and A-2-4 (0) in accordance with AASHTO soil classification and as SP-SC and SC-SM in the accordance with the USCS classification system. The soils were plasticity index was about 7 the liquid Limit was 23. The fines passing the #200 sieve measured at 11 and 14 in the tested samples. A summary of the laboratory test results are shown in Table 3.

It is anticipated that large boulders that may require the use of specialized fragmentation methods to remove from the excavation may be encounter in the footing excavations. Although ground water was not encountered during the investigation, water should be anticipated and the contractor should prepare the foundation in accordance with Section 208 of FP-03.

TABLE 3:- Laboratory Test Results for the Samples Taken at the East and West Abutment Locations

Test Sample TP-4 TP-5

Location Near Abutment #1(West Side) Near Abutment #2 (East Side) Station 48+65 49+15 Depth 3.0 to 5.0 feet 3.0 to 5.0 feet

Gradation AASHTO T 11 & 27 3” 100 100

1½” 97 98 1” 93 95 ¾” 90 92 ½” 80 85

3/8” 72 78 #4 (used to determine compaction) 58 64

#8 #10 48 54 #16 44 50 #30 #40 30 36 #50

#100 15 19 #200 11 14

Liquid Limit 23 23 Plasticity Index 7 5

AASHTO Classification A-2-4 (0) A-1-b (0) USCS Classification SP-SC SC-SM

BRIDGE FOUNDATION DESIGN

A prefabricated steel-girder bridge that is 16-feet wide and 50-long is proposed to be constructed between approximate Stations 48+50 and 50+ 00.

Specific site descriptions for foundation design are based on field observations, subsurface characterization information, laboratory testing, and may vary in material type and strength with distance away from the bridge site. In general, the subsurface can be characterized for design purposes as coarse gravel, cobbles and boulders in a coarse, medium dense, sand matrix from the surface to the depth of approximately 5 feet. Based on past experience with similar materials, conservative values for the medium dense granular sandy material was estimated from published data to have a unit weight (γ) of 125 pound per cubic foot, a friction angle (φ) of 32 degrees, and no cohesion.

The nominal service I limit state loading conditions for the new bridge were provided by CFL bridge personnel, and are estimated at 130 kips per abutment and the factored strength I limit state loads are estimated at 185 kips per abutment. Due to shallow scour conditions and relatively low bridge loads, the Bridge may be founded on a rectangular-reinforced concrete spread footing foundation. Semi-cantilever wingwalls may also be founded on the sandy/gravelly overburden soils at the same elevation as the bridge footing. The footings and wingwalls should be designed in accordance with the general guidelines specified in the AASHTO LRFD Standard Specifications. The wingwalls lengths should be sufficient to retain roadway embankment and provide scour and erosion protection. The estimated foundation elevation at each abutment is shown in Table 4.

TABLE 4: - Estimated Crossing Bridge Deck and Foundation Elevation at Proposed Roadway Center Line

BRIDGE ABUTMENT APROXIMATE STATION

ESTIMATED

BRIDGE DECK

ELEVATION,

ESTIMATED

FOUNDATION

ELEVATION

Abutment #1 48+65 ft 6,538.0 ft 6,530.9 ft Abutment #2 49+15ft 6,538.5 ft 6,531.4 ft

The spread footings width should be proportioned such that the supporting soil provides adequate bearing resistance, considering potential for geotechnical and structural failure and for tolerable settlement at the Strength I and Service I Limit States, respectively. For foundations on medium dense sands, a presumptive nominal bearing resistance (qn) at the Service I Limit State of 4,000 pounds per square foot (AASHTO Table C10.6.2.6.1-1) is recommended to prevent shear failure of the supporting materials and to limit vertical deformation to one-inch. The designed footing foundation must also accommodate external forces including sliding and overturning (eccentric loading) applied to the footing. Lateral resistance due to sliding should be calculated using a coefficient of friction between the sandy soil and a cast-in place concrete footing is 0.6 with a shear resistance factor of 0.8 for LRFD design.

To evaluate the bearing resistance (qn) available for the Strength I and Extreme Event I limit states, and the settlement limited nominal bearing resistance (qserv) for one-inch of settlement for various effective footing widths the resistance factors for each limit state are provided Table 5.

TABLE 5 - Resistance Factors φ for each Limit State

A minimum footing width of four feet is recommended to account for site variation. Design the footing to bear below the frost depth of 36 inches (below adjacent grade within 10-foot radius) for the project site.

This is required to prevent heaving and to prevent settlement due to loss of shear strength and stiffness from thawing. Frost heave is unlikely due to the low amount of silts in the overburden soils.

The foundation should be constructed below the scour depth as reported by CFL Hydraulics Section, and prepared and compacted in accordance with Section 208 of the FP-03. Boulders equal or greater than 4 feet in diameter may be encountered during excavation. Boulders encountered at the footing elevation should be removed to a depth of at least six-inches below the foundation elevation to provide a relatively level pad for constructing the foundation. Minor concrete meeting section 601 of the FP-03 should be used in boulder prone areas to fill any voids and to provide a level pad for footing. In areas where the

LIMIT STATE BEARING SHEAR RESISTANCE TO

SLIDING

PASSIVE PRESSURE RESISTANCE

TO SLIDING

Strength 0.45 0.80 0.50 Service and

Extreme Event 1.00 1.00 1.00 footing bears on soil and bedrock along the entire footing, it is recommended that a minimum of six inches of rock be sub-excavated and replaced with foundation fill compacted to a minimum of 95 percent relative compaction (ASTM D1557) to avoid adverse differential settlement. Blasting of boulders or bedrock is not recommended but if rock breakage is required to facilitate rock removal, rock fragmentation (breakage) should only be accomplished using expansive grouts or equivalent methods. If rock fragmentation result in overbreak and fracturing that extend beyond the foundation elevation it may reduce the resistance to scour in the zone of overbreak. Fracture filling with lean concrete may be required if these techniques were used for foundation excavation.

BRIDGE ABUTMENT DESIGN

The semi-gravity abutment walls should be designed to resist lateral earth pressures and other applicable lateral loads in accordance with the AASHTO LRFD Specifications. Lateral earth pressure acting on the back of the abutment wall is dependent on the following factors: (1) backfill material strength properties, including soil type and gradation, unit weight, moisture content, and friction angle, (2) backfill drainage,

(3) soil-structure interaction and, (4) compaction induced pressure and other surcharges.

Backfill material strength properties are estimated based on the field investigations and presumptive values and are presented in Table 6.

TABLE 6: - Recommended Wall Backfill Parameters for Design.

MATERIAL

DESCRIPTION

UNIT WEIGHT

ɣ, pcf

COHESION

c, psf

FRICTION ANGLE

φ, deg

Structural Backfill 120 0 35 Unclassified Borrow 125 0 32

Unbalanced groundwater behind walls adds significant lateral pressure and should be avoided by using permeable backfill and assuring a free draining gravity outlet for water. Backfill in the immediate vicinity of the retaining structure should consist of structural backfill in accordance with subsection 704.10 of FP-

03. Unclassified borrow is typically used in the retained earth zone behind the wall footing.

Rigid, absolutely retained walls that can tolerate little or no movement should be designed for the at-rest earth pressure (Ko) by an equivalent-fluid soil unit weight of 57 pounds per cubic foot for unclassified backfill material. That value is based on a compacted moist unit weight of 125 pounds per cubic foot and an effective friction angle of 32 degrees for leveled, free draining structural backfill, resulting in an at-rest earth pressure coefficient of 0.45.

Retaining walls that can tolerate sufficient rotation at the top of the wall or lateral translation equal to or greater than 0.002 times the wall height should be designed for an active lateral earth pressure (Ka). The Rankine's lateral-active-earth pressure represented by an equivalent-fluid pressure of 37 pounds per cubic foot for leveled, free draining backfill can be assumed as the lateral earth pressure acting at a distance of one-third the wall height above the wall base.

The Rankine's method can be used if the “soil-structure interaction” is neglected. To account for the soil-structure interaction with a friction angle valued at 2/3 times the backfill effective friction angle (0.67 φ´), the Coulomb's method may be used for calculating the lateral-active- earth pressure.

In addition to the lateral earth pressure, retaining structures should be designed considering additional pressures due to traffic live-load surcharge equivalent to 2 feet of soil (250 pounds per cubic foot) as described in Article 3.11.6.4 of AASHTO LRFD (2010).

When passive lateral earth pressures are applicable, a wall movement of 0.02 times the wall height at the top of the wall is necessary to reach passive earth pressure. Procedures outlined in Appendix A11 of the AASHTO LRFD should be used during design to determine forces exerted on the walls.

Lateral loads imposed on the wall due to lateral earth pressure and horizontal seismic forces may be resisted by the friction developed between the cast-in-place concrete footing and the supporting soils with a resistance factor of 0.8. The nominal sliding resistance for concrete footings cast on cohesionless soil (native soil, foundation fill and decomposed/weathered bedrock) is 0.65.

EARTHWORK RECOMMENDATIONS

It is anticipated that the planned fills will be mostly constructed using locally available materials derived from the project excavations. A shrink/swell factor of 1.0 is recommended for the materials at the site for material estimating purposes. Portable crusher and a screening plant to remove fines and reduce over-sized rock within excavated materials to meet the select granular backfill specifications may be used if cleared by the environmental documents. If materials are cleaned by removing fines for use as backfilling, an approximate shrink value of five percent is recommended as an earthwork factor to account for the volume change from excavation to the compacted embankment. The construction techniques used in the field will have significant influence on the actual shrink/swell of the excavated materials. The in-place embankment densities will also be influenced by the gradation and screening of the fill materials, staging and placement, and method of compaction. In general excavation of large boulders (greater than 6 feet in diameter) is not recommended except for bridge footing preparations. Top soil is approximated to vary between zero and 6 inches in thickness (averaging three-inches for estimation purposes).

Structural foundation excavations should be in accordance with Section 208 of FP-03. Soils and boulders should be excavated to the limits of the footing elevations and compacted in accordance with Section

204.11 of FP-03. Excavate any unstable materials below foundation elevation and replace with foundation fill and compact in lifts not to exceed 6 inches in height. Dewater the foundation and assure that stable soils exist before forming the bridge footings.

The appropriate Occupational Safety and Health Administration Regulations for temporary excavation stability must be followed during construction. The overburden materials to be excavated should be classified by the contractor based on a field site visit prior to bidding. Any zone identified for excavation at a steeper angle than 45 degrees should be approved by the CO before excavating.

If you have any questions please do not hesitate to call me at 720 963-3521.

______________________________________________ Date_________________ Khamis Y. Haramy Lead Geotechnical Engineer, CFLHD Phone: 720.963.3521

figures for memo.