160603-Truckee_River_Bridge_and_Retaining_Walls_Geotechnical_Report-final.pdf

Geotechnical Report Central Federal Lands Highway Division

CA SR 89/SR 28 Intersection Area Bridge No. 19-0205, Placer County, Route 89, PM 8.2

Caltrans EA: 03-3A7601 Caltrans PIN: 0300000452

Truckee River Bridge Tahoe City, California

May 20, 2016

Submitted To:

Mr. Bill Lang

CH2M Hill, Inc.

9191 South Jamaica Street

Englewood, CO 80112

By:

Shannon & Wilson, Inc.

400 N 34th Street, Suite 100 Seattle, Washington 98103

23-1-01373-205 i

TABLE OF CONTENTS

Page

1.0 INTRODUCTION

2.0 PROJECT AND SITE DESCRIPTION

3.0 FIELD EXPLORATIONS AND LABORATORY TEST RESULTS

3.1 Subsurface Explorations

3.2 Laboratory Test Results

4.0 SUBSURFACE CONDITIONS

4.1 Abutment 1 - West Shore

4.2 Abutment 2 - East Shore

4.3 Groundwater

4.4 Scour

5.0 GEOLOGY AND GEOLOGIC HAZARDS

5.1 Regional Geology

5.2 Geologic Hazards

5.2.1 Seismic Ground Motion

5.2.2 Liquefaction

5.2.3 Slope Stability

5.2.4 Lateral Spreading and Earthquake-induced Settlement

5.2.5 Fault Hazard

6.0 CONCLUSIONS AND RECOMMENDATIONS

6.1 Abutment Foundation Design

6.2 Abutment Wall Design

6.3 Design of Other Retaining Walls

6.3.1 Lateral Pressures

6.3.2 Bearing Resistance

6.3.3 Sliding Resistance

6.3.4 Wall Drainage

6.4 Pavements

7.0 CONSTRUCTION AND MATERIALS SPECIFICATIONS

7.1 Site Preparation

7.2 Backfill Material, Placement, and Compaction

7.3 Temporary Groundwater Control

7.4 General Excavation, Temporary Cut and Permanent Slopes

7.5 Utilities

8.0 PLAN REVIEW AND CONSTRUCTION OBSERVATION

9.0 LIMITATIONS

10.0 REFERENCES

ii

TABLES

1 Scour Data 2 Design Response Spectrum 3 Spread Footing Data Table 4 Recommended Soil Parameters for Abutment Design 5 Recommended Soil Parameters for Rockery Walls 6 Recommended Soil Parameters for Cast-in-Place Walls 7 Recommended Parameters for Rockery and Cast-in-Place Sliding

FIGURES

1 Vicinity Map 2 Site and Exploration Overview 3 Truckee River Bridge Site and Exploration Plan 4 Fanny Bridge Site and Exploration Plan 5 Design Response Spectrum 6 Global Stability Analyses, Static 7 Global Stability Analyses, Seismic 8 Global Stability Analyses, Post-Seismic 9 Factored Bearing Resistance Abutment Footings Rectangular Footing, L/B=1.85 10 Factored Bearing Resistance Rockery and C.I.P. Retaining Walls Rectangular

Footing, L/B 10

APPENDICES

A Previous Subsurface Explorations B Laboratory Test Results Performed by Wood Rodgers C Wood Rodgers ReMi Geophysical Survey D Example Liquefaction Calculation E Response to Caltrans Comments F Important Information About Your Geotechnical Report

GEOTECHNICAL REPORT

CENTRAL FEDERAL LANDS HIGHWAY DIVISION

CA SR 89/SR 28 INTERSECTION AREA

TRUCKEE RIVER BRIDGE AND FANNY BRIDGE

BRIDGE NO. 19-0205, PLACER COUNTY, ROUTE 89, PM 8.2

CALTRANS EA: 03-3A7601, CALTRANS PIN: 0300000452

TAHOE CITY, CALIFORNIA

1.0 INTRODUCTION

This geotechnical report provides geotechnical recommendations towards delivery of a final set of plans, specifications and estimate for the Federal Highway Administration, Central Federal Lands Highway Division (CFLHD) California State Route (SR) 89/ SR 28 intersection area near Tahoe City, California (Figure 1). Our work was completed in general accordance with our scope of work and purchase order agreement with CH2M Hill, Inc. (CH2M Hill) dated August 17, 2015 and modified September 30, 2015.

This report presents recommendations only for the design and construction of the Truckee River Bridge and includes subsurface explorations, laboratory testing, and design and construction recommendations.

Figures 2 through 4 provide the location of the borings that were performed for the Fanny Bridge and project-wide pavement design. Recommendations for design and construction of the Fanny River Bridge and project-wide pavement recommendations are presented in a separate report.

2.0 PROJECT AND SITE DESCRIPTION

The SR 89 route is currently the only transportation route along Lake Tahoe’s west shore for visitors making connections from the national highway system, US 50 and Interstate 80. The two lanes SR 89 along the west shore also provides the only access to over 25,000 acres of federal lands managed by the (United States Forest Service) USFS – Lake Tahoe Basin Management Unit. The historic Fanny Bridge is narrow at 2 lanes and does not provide adequate access for the 22,000 vehicles per day and 400 bike and pedestrians per hour during peak summer months with visitors accessing the USFs and State Parks recreational areas. Backups in Tahoe City at Fanny Bridge have extended over 2 miles south down SR 89 equating to 2 hour or more delays.

Congestion issues also pose safety risks in emergency response time delays. The proposed improvements would enhance the Tahoe City and west shore community with improved access to federal lands and enhancing bike and pedestrian facilities removing bike/pedestrian traffic from the narrow SR 89 shoulders.

Project features as shown in Figure 2 include:

A new bridge over the Truckee River and associated approaches Replacement of existing Fanny River bridge New bike paths under the new Truckee River bridge that tie into existing trails Construction of east, west, and WYE roundabouts Pavements associated with the proposed bridges and roundabouts Retaining walls (i.e. rockery, Type 1 cast-in-place, etc.)

The proposed Truckee River Bridge is a key component of the project to alleviate traffic congestion at the intersection of SR 89 and SR 28. The proposed bridge would cross both the Truckee River and a regional trail. The proposed Truckee River Bridge is approximately 1,600 feet down river of the Fanny Bridge (Figure 2).

3.0 FIELD EXPLORATIONS AND LABORATORY TEST RESULTS

3.1 Subsurface Explorations

Subsurface information for the Truckee River Bridge was provided by CFLHD. Seven borings ranging between 61.5 and 81.5 feet deep with two additional borings outside the proposed bridge alignment were performed by Wood Rodgers under contract to CFLHD. A geophysical survey consisting of five Refraction microtremor (ReMi) profiles was performed by Wood Rodgers.

The ReMi geophysical measurements provide a shear wave velocity profile that is developed using passive measurement of ambient ground vibrations. The Wood Rodgers boring logs and laboratory test results are presented in Appendix A and B, respectively. The ReMi data report is presented in Appendix C. The locations of the borings and ReMi survey lines are shown in Figures 2 through 4. We understand that Placer County permits were obtained to advance the borings and that the borings backfilled in accordance with Placer County regulations. The Wood Rodgers borings do not provide a measurement of SPT hammer energy. Per AASHTO specifications, the hammer energy is not required to be measured and may be assumed to be 80 percent.

As part of this evaluation, Shannon & Wilson advanced cone penetration tests (CPT) at the site.

The objective of the CPTs was to evaluate the areal extent of the silt layer encountered in the previous Wood Rodgers explorations. The CPT rig was equipped with a hydraulic hammer to advance the exploration in the event that the CPT could not penetrate very dense soil. Details of the subsurface conditions encountered by the CPTs are provided below. The CPT locations are shown in Figures 2 and 3.

Three CPTs were attempted on the east shore to evaluate the areal extent of the silt layer observed in the Wood Rodgers borings and determine the thickness of the silt layer. CPT-E1 and CPT-E2 were pre-drilled through very dense sandy gravel to a depth of about 17 feet. At this depth, the fluid in the hole was not sufficient to resist the upward pressure and about 3 to 5 feet of heave occurred in the two explorations. These two explorations were not able to advance beyond this depth. The soil that heaved consisted of poorly graded sandy gravel. CPT-E3 was pre-drilled through very dense sandy gravel to a depth of about 19 feet and the CPT was able to advance to a depth of 22 feet before encountered refusal conditions.

Additionally, three CPTs were attempted on the west shore to evaluate the areal extent of the silt layer observed in the Wood Rodgers borings and determine the thickness of the silt layer. CPT- W1 was drilled 9 feet through very dense silty gravel until 2 to 3 feet of heave occurred in the exploration and the exploration could not continue beyond this depth. CPT-W2 was drilled 19 feet through relatively dense to very dense silty sand and gravel and did not encounter a loose low plasticity, sandy silt and silty sand similar to that encountered in the previous explorations and at a depth of 19 feet the exploration was terminated. CPT-W3 was drilled 24 feet through very dense silty sand with gravel until 4 to 5 feet of heave occurred in the exploration and the exploration could not continue beyond this depth. The data report from the CPT contractor is presented in Appendix A.

3.2 Laboratory Test Results

Laboratory tests performed on samples from the Wood Rodgers borings consist of moisture content, grain-size distributions, Atterberg limits, and direct shear tests. The laboratory test results are reproduced in Appendix B. The engineering soil properties we developed for the site consider many factors including the Wood Rodgers laboratory test results, the in-situ standard penetration tests, ReMi geophysical testing, and appropriate engineering property correlations.

4.0 SUBSURFACE CONDITIONS

Based on the Wood Rodgers borings, the subsurface conditions at the Truckee River Bridge typically consisted of granular soils with variable silt content. The general subsurface conditions observed at the Truckee River Bridge abutments are presented in the following sections.

4.1 Abutment 1 - West Shore

On the west shore of the river, the idealized subsurface profile consists of the following soil layers:

A surficial layer that is 10 to 15 feet thick containing loose to dense, silty sand with gravel and clayey sand with gravel

Underlain by 2 1/2 to 5 feet of very loose to loose, low plasticity, sandy silt and silty sand. The plasticity index if this layer is about 4. The SPT N-values range between 5 blows per foot for the sandy silt to 27 blows per foot for the silty sand. While this layer was observed by Wood Rodgers, it was not apparent during the CPT investigation and may be discontinuous.

Dense to very dense, gravelly sand to the remaining depth of the explorations.

4.2 Abutment 2 - East Shore

On the east shore of the Truckee River, the generalized subsurface profile consists of the following soil layers:

A surficial layer that is 10 to 12 feet thick containing loose to dense, silty sand with gravel

Underlain by 15 feet of medium dense to very dense, silty sand to silty gravel near Abutment 1.

5 to 7 feet of very loose to loose, low plasticity, sandy silt. The plasticity index if this layer is approximately 2 to 7. Standard penetration test (SPT) N-values range from 3 to 13 blows per foot.

25 to 50 feet of high plasticity, elastic silt near borings B-1 and B-2. Borings B-1 and B- 2 are approximately 120 to 200 feet east of Abutment 2, respectively.

Dense to very dense, gravelly sand to the remaining depth of the explorations. Basalt was encountered in the bottom of boring B-1.

4.3 Groundwater

The groundwater level was observed in the Wood Rodgers explorations at a depth of about 5 to 10 feet on both sides of the river. For evaluation purposes, the groundwater elevation was assumed to be equal to the Truckee River which is about 6,220 feet. We understand that the river elevation varies depending on the flow allowed by the Bureau of Reclamation.

4.4 Scour

The following Table 1 presents the scour data for the bridge abutments. The scour data in Table 1 was obtained from the project hydraulic report and project drawings prepared by others.

TABLE 1

SCOUR DATA

100-year

Support No.

Long Term Scour (degradation, ft)

Short-Term Scour (ft) General Contraction Local

Depth Elevation Depth Elevation Depth Elevation Abutment 1 0.00 6222.86 0.74 6222.12 5.17 6217.69 Abutment 2 0.00 6222.04 1.35 6220.69 4.64 6217.40

500-year

Support No.

Long Term Scour (degradation, ft)

Short-Term Scour (ft) General Contraction Local

Depth Elevation Depth Elevation Depth Elevation Abutment 1 0.00 6222.86 0.87 6221.99 6.55 6216.31 Abutment 2 0.00 6222.04 1.48 6220.56 5.44 6216.60

Notes: 1) Per NCHRP methodology for local abutment scour, general contraction scour values are not additive, but instead are already included.

5.0 GEOLOGY AND GEOLOGIC HAZARDS

5.1 Regional Geology

The project area is located near the western shoreline of Lake Tahoe along the eastern extents of the Sierra Nevada Mountain Range, which comprises a tilted fault block approximately 400 miles long. Mesozoic granite rock forms the Sierra Nevada Batholith, which underlies the Tahoe Basin (Hyne, 1972). The granitic rock along the perimeters of Lake Tahoe is extensively weathered due to extreme water fluctuations from recurrent glacial damming in the Pleistocene.

Tertiary volcanic rocks overly the granites and includes basalt, andesite, latite, trachytes, tuff, and mudflows. These tertiary units are frequently observed at the ground surface (Grose, 1985).

Other deposits that are often observed locally are alluvium, colluvium, and glacial materials.

5.2 Geologic Hazards

5.2.1 Seismic Ground Motion

The seismic analysis is based on ground motion levels, performance criteria, and methods as described in AASHTO (6th Edition, 2012) and the Caltrans Amendments to AASHTO, dated January 2014. The site class is determined using standard penetration test (SPT) N-values and/or shear wave velocities measured in the upper 100 feet of the soil profile. The soil classification at the Truckee River Bridge location was derived based on the Wood Rodgers explorations completed near the proposed bridge. The average SPT N-value for the upper 100 feet of boring B-1, nearest the bridge, is 35 blows per foot. Additionally, a geophysical survey was performed by Wood Rodgers using ReMi to estimate the shear wave velocity near the bridge abutments.

The results of the study are provided in Appendix C. According to the survey results, the average shear wave velocity in the upper 100 feet (Vs30) is approximately 900 to 1,180 feet per second (275 to 360 meters per second). Using the AASHTO site class criteria, the subsurface SPT N-values and shear wave velocities near the Truckee River Bridge indicate Site Class D should be used in design.

The design response spectrum corresponds to a 975-year return period ground motion and was determined for the project site using the Caltrans ARS Online application (Version 2.3.06), which considers both deterministic and probabilistic seismic hazard methods. The design response spectrum is based on the larger of deterministic and probabilistic spectral accelerations.

The design response spectrum incorporates soil effects using a Vs30 equal to 900 feet per second for the project site. The design response spectrum is presented in Figure 5 and numerical values are presented in Table 2.

5.2.2 Liquefaction

Based on AASHTO and the Caltrans Geotechnical Manual, we estimated the liquefaction potential at the Truckee River Bridge for the 975-year return period ground motion. The liquefaction analysis described below is based on Youd et al. (2001) and provides the consensus of the profession up to the year 2000. More recent liquefaction analysis methods are available and could provide a slightly different estimate of liquefaction potential; however, the Youd et al.

(2001) method is typically considered to be the benchmark method for other liquefaction analysis methods and is recommended by AASHTO and the Caltrans Geotechnical Manual (2014).

For liquefaction potential of low plasticity silt we considered the methods recommended by Idriss and Boulanger (2008) and Bray and Sancio (2006). The Caltrans Geotechnical Manual (2014) recommends the Bray and Sancio (2006) method and both methods are included in AASHTO.These methods are a liquefaction screening criteria based on plasticity index and/or natural water content. Both methods indicate the low plasticity silt layer is susceptible to liquefaction and should be considered in the liquefaction potential evaluation using a method such as Youd et al. (2001).

The peak ground acceleration for this site is 0.43g. Based on the USGS seismic hazard deaggregation for the site the magnitude appropriate for the 975-year return period ground motion and corresponding liquefaction analysis is about 6.8. This magnitude is associated with the North and West Tahoe faults that are about 3 to 6 miles (5 to 9 kilometers) from the site.

TABLE 2

DESIGN RESPONSE SPECTRUM

Period (seconds) Spectral Acceleration (g)

0.01 0.43

0.05 0.66

0.10 0.79

0.15 0.90

0.20 0.99

0.25 0.97

0.30 0.95

0.40 0.86

0.50 0.80

0.60 0.74

0.70 0.70

0.85 0.64

1.00 0.59

1.20 0.51

1.50 0.43

2.00 0.32

3.00 0.20

4.00 0.14

5.00 0.11

Notes: 1) Spectral accelerations are the larger of the deterministic and probabilistic values obtained from Caltrans ARS Online.

2) Spectral values include near fault factors as incorporated by Caltrans ARS Online.

3) Spectral values are based on Vs30 = 900 ft/s.

Liquefaction of loose, saturated, cohesionless soils due to seismic loading has been studied extensively over the past 40 years, resulting in both laboratory and field procedures to evaluate liquefaction potential. The analyses presented in this report are derived in part from the recommendations presented in Seed and Idriss (1971), NCEER (1997), Youd et al (2001), Robertson & Wride (1998), and Robertson (2009). Liquefaction potential is reported in terms of a factor of safety (FS) representing the ratio of soil resistance to seismic demand. Soil that has a FS against liquefaction greater than 1 is suitable for the 975-year return period ground motion.

DemandSeismic ResistanceSoilFS

The soil’s resistance to liquefaction is termed cyclic resistance ratio (CRR). For the site specific SPT data, the CRR is estimated using the empirical relationship by Youd et al (2001).

This empirical relationship relates N1,60,CS to CRR for a moment magnitude of 7.5 at a vertical effective stress of 1 atmosphere (atm), or CRR7.5, 1atm. The field blow count values, N, were corrected for energy, overburden stress, and fines content per the procedures in Youd et al (2001).

Hammer energy for the auto-hammer used during drilling was assumed to be 80 percent according to AASHTO. The cone penetration test data was interpreted utilizing the procedures described in Robertson (2009).

Since the magnitude associated with the 975-year return period earthquake is not equal to 7.5, the CRR7.5 was adjusted by means of a magnitude scaling factor (MSF) as follows:

Where the MSF recommended in TRB (1999), also referred to as the “lower bound” in

Youd et al, was chosen for this project and is calculated as:

= 10 ..

The seismic demand is termed cyclic stress ratio (CSR) and is expressed as the ratio of average equivalent shear stress to the initial vertical effective overburden stress. The cyclic stress ratio of soil at depth in a soil deposit is estimated by the following equation (Seed and Idriss, 1971):

d vo vo vo r g aCSR ' max

' 65.0 where amax is the peak ground acceleration generated at the ground surface, g is the acceleration of gravity, vo and 'vo are the total and effective vertical overburden stresses, respectively, and rd is a stress reduction coefficient. The stress reduction coefficient adjusts the ground surface PGA for the response of the soil profile as a function of depth. The relationship of rd as reported by Liao and Whitman (1986) and Youd et al (2001), which represents the average values reported by Seed and Idriss (1971), was used to calculate the stress reduction coefficient. Examples of liquefaction factor of safety calculations performed for our analysis are presented in Appendix D.

The soil encountered in borings WR1, WR2, WR4, WR5, B1, B2, and B3 and CPT-E3 were evaluated for liquefaction potential for the design ground motion. The results of this evaluation indicate that the 5 to 7 foot thick layer of very loose to loose, low plasticity, sandy silt would liquefy on the east shore. This low plasticity, sandy silt layer is about 25 to 27 feet below the existing east shore ground surface. The liquefaction potential of the low plasticity sandy silt and silty sand includes consideration of the behavior of low plasticity silt based on the Idriss and

Boulanger (2008) and Bray and Sancio (2006) criteria. In boring WR5 near the east shore and the Tahoe River trail, two samples at a depth of about 35 feet are potentially liquefiable during the design ground motion.

On the west shore at a depth of about 10 to 15 feet, the 2 1/2 to 5 feet of very loose to loose, low plasticity, sandy silt and silty sand is potentially liquefiable; however the extent of this material was not confirmed during the CPT exploration program. The liquefaction potential of the low plasticity sandy silt and silty sand includes consideration of the behavior of low plasticity silt based on the Idriss and Boulanger (2008) and Bray and Sancio (2006) criteria.

5.2.3 Slope Stability

Slope stability evaluations were performed to assess static, seismic, and post-liquefied stability, of the Truckee River Bridge approach embankments. The slope stability cases considered include: (1) static, (2) seismic (static soil properties and a seismic coefficient), and (3) post-seismic (liquefied soil properties as appropriate and no seismic coefficient).

The soil parameters for these stability evaluations were estimated based on field samples, laboratory tests, empirical relationships and our experience. Using the empirical relationships developed by Olsen and Stark (2002), the normalized residual shear strength (Su/ v’) of the liquefied sand is expected to be approximately equal to 0.1, where Su is the undrained residual shear strength and v’ is the effective overburden stress. The results of the static, seismic, and post-seismic evaluations are presented in Figures 6 through 8 and summarized below:

For static loading the minimum factors of safety on the east and west shore are greater than 1.5 and are suitable (Figure 6).

For seismic loading a pseudo-static inertial acceleration of one-third (Caltrans Bridge Design Specifications) the ground surface peak ground acceleration equal to 0.15g was assumed. The lowest factor of safety for the east and west shores is shown in Figure 7and is larger than the minimum 1.1 required by the Caltrans Geotechnical Manual (2014).

For the Abutment 1 and 2 post-seismic analyses the inertial loading due to the design ground motion is assumed to have ended and the low plasticity sandy silt and silty sand is assumed to have liquefied. The block failure surfaces evaluated indicate factors of safety of about 1.0 as shown in Figure 8. This factor-of-safety assumes the low plasticity sandy silt and silty sand layer extends below the Truckee River. If the low plasticity sandy silt and silty sand layer does not extend below the Truckee River then the factor of safety would be larger than 1.0.

As noted above, our analysis indicates that the potentially liquefiable low plasticity sandy silt and silty sand is approximately 12 feet deep beneath Abutment 1 at the west shore and at least 25 feet deep beneath Abutment 2 on the east shore. In general, the soil layer surrounding low plasticity sandy silt and silty sand is generally dense which would provide confinement and limit the potential for the silt layer to strain and generate excess pore pressure (i.e. liquefy).

Additionally, these low plasticity sandy silt and silty sand layers were not observed to be laterally continuous across the site.

5.2.4 Lateral Spreading and Earthquake-induced Settlement

Lateral spreading is not expected at the abutments due to the level of ground shaking, the limited thickness of liquefaction, and the potential that the soil layers at the site are discontinuous.

Using the results of the liquefaction analysis presented above, we estimate that post-seismic abutment and approach settlement could be about 1 to 2 inches based on the Tokimatsu and Seed (1987) method. Given the relatively small estimated liquefaction-induced settlement, assumptions regarding the distribution of settlement at the site are beyond the accuracy of the empirical methods.

5.2.5 Fault Hazard

There are three nearby faults that are classified as Quaternary Class A or Class B faults.

The Incline Village fault is less than 15,000 years old and is located approximately 9 miles to the east of the site. The North Tahoe Fault is less than 15,000 years old and is located approximately 5 miles east of the site. The West Tahoe-Dollar Point fault is less than 130,000 years old and is located approximately 3 miles from the site. The 1972 Alquist-Priolo Earthquake Zoning Fault Act has not designated the locale as an Earthquake Fault Zone (EFZ).

6.0 CONCLUSIONS AND RECOMMENDATIONS

The proposed Truckee River Bridge is a 153-foot long and 44-feet wide single span bridge and supported by 17-foot wide box culverts at both ends. The clear span distance is 119 feet.

Abutments 1 and 2 are on the west and east sides of the river, respectively.

The ground surface elevation at the Truckee River Bridge varies between 6,220 and 6,230 feet.

The bridge deck elevation is approximately 6,238 feet. Construction of the bridge will require about 8 to 18 feet of fill at the abutments. The 100- and 500-year high water elevations are 6,224.8 and 6,225.8 feet, respectively.

6.1 Abutment Foundation Design

Based on the borings and discussions with the structural engineer, the Truckee River Bridge will be founded on box culvert type spread footings that bear in the medium dense to dense, silty sand with gravel. The bottom of footing elevation at Abutment 1 and 2 is 6,220 and 6,219.25, respectively.

The bearing resistance, including resistance factors, for the length to width ratio that corresponds to the proposed box culvert dimensions is provided in Figure 9. For our bearing resistance analysis, we used elastic deformation methods to calculate the load at which an allowable settlement is reached. For the service limit bearing resistance that corresponds to 1 inch of static settlement the subgrade modulus is 50 pounds per cubic inch. For sliding resistance determination, we recommend that an internal friction angle, f (AASHTO Section 10.6.3.4), equal to 34 degrees be used between cast-in place concrete and compacted sand and gravel. For the strength limit state, a (AASHTO Table 10.5.5.2.2-1) should be applied to the nominal coefficient of sliding resistance value. For the extreme limit state, a

1.0 (AASHTO Section 10.5.5.3.3) should be applied to the nominal coefficient of sliding resistance value. The spread footing data table is presented in Table 3.

TABLE 3

SPREAD FOOTING DATA TABLE

Support No.

Service Permissible Net Contact Stress

(Settlement) (ksf)

Strength/Construction Factored Gross

Nominal Bearing Resistance b=0.45 (ksf)

Strength/Constr uction Factored Gross Nominal

Bearing Resistance b=1.0 (ksf)

Abut 1 4.3 20 44.3 Abut 2 4.3 20 44.3

The service limit state bearing resistance is provided in Figure 9 as a function of the total elastic settlement anticipated. Differential settlement is estimated to be approximately ½ of the total settlement. The settlement is expected to occur as the load is applied.

The bottom of bridge spread footing foundations should extend a minimum of 2 feet below lowest adjacent finished grade for frost protection. To reduce the potential for differential settlement, the footing subgrade should be proof-rolled and recompacted as described in Section 7 below.

6.2 Abutment Wall Design

Based on the project plans and subsurface conditions from the site investigations, 10 to 15-foot-tall retaining walls will be required at the bridge abutment to support the proposed grades. We understand the abutment walls will be free to deflect and can therefore be designed for the active condition.

We recommend that fill material used to backfill the retaining walls consist of free draining, sand and gravel Structural Backfill conforming to Section 704.04 of the 2014 Federal Highways Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects (FP-14). Based on the use of Structure Backfill, the parameters shown in Table 4 are recommended for the design of abutment walls.

The parameters provided for the retaining walls assume a free-draining backfill condition. As such, it will be important to control surface water and reduce its potential to infiltrate into the ground around the retaining walls. Surface water behind the wall should not be allowed to discharge directly into the wall backfill materials. In addition, water should not be allowed to discharge or pond around walls. Weep holes should be incorporated into the retaining walls to provide for drainage of backfill should groundwater elevations rise. A geosynthetic filter fabric should be placed behind weep holes to reduce the potential for soil loss.

TABLE 4

RECOMMENDED SOIL PARAMETERS FOR ABUTMENT WALL DESIGN

Design Parameter Recommended Value Total Unit Weight of Backfill (pcf) 120 Friction Angle (degrees) 34 Cohesion 0 Equivalent Fluid Density for Active Retaining Conditions (AASHTO Section 3.11.5.5) (pcf)

Horizontal 34

3H:1V backfill slope 42

2H:1V backfill slope 50 Active Earth Pressure Coefficient with Earthquake Effect (Kae) for Horizontal Backfill (AASHTO Section A11.3.1)

0.37

Note: H:V = horizontal:vertical

a) For the strength limit state, a resistance factor ( ) of 0.8 should be applied to the nominal coefficient of sliding resistance value (AASHTO Table 10.5.5.2.2-1).

(AASHTO Section 10.5.5.3.3) should be applied to the nominal coefficient of sliding resistance value.

b) Load factors have not been included in any of the values in this table.

In accordance with the Caltrans Bridge Design Specifications (2004) and the Caltrans Geotechnical Manual (2014), the horizontal coefficient (kh) is equal to one-third peak ground acceleration which is equal to about 0.15g for the project site. As stated in AASHTO 11.6.5.3, the seismic active earth pressure should be distributed using the same distribution as the static active earth pressure (i.e. triangular distribution increasing with depth) but the resultant force should not be applied less than H/3 above the wall base, where H equals the wall height. We recommend the procedure in AASHTO Section 11 Appendix A11 be used to compute the resultant force location when combining active and dynamic earth pressure (i.e. applied at 0.4H to 0.5H). We also note that equation A11.3.1-1contains an error in AASHTO 2012 and the correct equation is provided in AASHTO 2014.

The post-seismic slope stability factor of safety is 1.0 and is the result of an approximately 5 foot thick layer of very loose to loose, low plasticity, sandy silt that extends beneath the abutments based on the existing explorations. Because, the post-seismic slope stability factor of safety is

1.0 some post-seismic abutment and approach settlement could occur following the design ground motion. We estimate that post-seismic abutment and approach settlement could be about 1 to 2 inches.

The static settlement of the proposed 10- to 15-foot high embankments could be 1 to 2 inches and this settlement should occur as the embankment load is applied. The estimated static settlement is based on empirical correlations between soil modulus and SPT blow count.

coefficient of sliding resistance value (AASHTO Table 10.5.5.2.2-1). For the extreme limit nominal coefficient of sliding resistance value. Load factors have not been included in the Table 4 values.

6.3 Design of Other Retaining Walls

In addition to the abutment walls, 4 to 12-foot tall rockery and cast-in-place retaining walls will be required to support the roadway grades elsewhere on the alignment. This section provides geotechnical parameters for the design of the rockery walls. The parameters were developed in accordance with AASHTO and the Rockery Design and Construction Guidelines, Publication No. FHWA-CFL/TD-06-006, November 2006 (Rockery Guidelines).

Rockery walls are planned in both cut and fill areas of the alignment. Fill walls will support backfill placed during construction and cut walls will support the existing earth materials. In both cases, the Rockery Guidelines recommend a minimum of 12 inches of Crushed Rock

Backdrain should be placed directly behind the wall to avoid the buildup of pore water pressure (see Table 9 of the Rockery Guidelines). Cast-in-place walls should be backfilled with a free draining, sand and gravel Structural Backfill conforming to Section 704.04 of FP-14. Rockery and cast-in-place retaining walls should extend a minimum of 2 feet below lowest adjacent finished grade for frost protection.

6.3.1 Lateral Pressures

Both rockery and cast-in-place retaining walls may be designed for active conditions.

We recommend that AASHTO equation 3.11.5.3-1 be used for static active design of rockery and cast-in-place walls. We recommend that AASHTO equation A11.3.1-1 be used for seismic active design of rockery and cast-in-place walls. These AASHTO equations allow for variation of ). Based on the use of Crushed Rock Backdrain for rockery walls and Structure Backfill for cast-in-place walls, the parameters shown in Tables 5 and 6 are recommended for the design of rockery and cast-in-place walls, respectively. As stated in AASHTO 11.6.5.3, the seismic active earth pressure should be distributed using the same distribution as the static active earth pressure (i.e. triangular distribution increasing with depth). However the resultant total active force should not be applied less than H/3 above the wall base, where H equals the wall height. We recommend the procedure in AASHTO Section 11 Appendix A11 be used to compute the resultant force location when combining active and dynamic earth pressure (i.e. applied at 0.4H to 0.5H). We also note that equation A11.3.1-1contains an error in AASHTO 2012 and the correct equation is provided in AASHTO 2014.

As previously discussed, the horizontal coefficient (kh) is equal to one-third peak ground acceleration, or 0.15g. This value assumes the rockery and cast-in-place walls can tolerate displacements of 1 to 2 inches or more during the design seismic ground motion.

The parameters provided for the retaining walls assume a free-draining backfill condition. As such, it will be important to control surface water and reduce its potential to infiltrate into the ground around the retaining walls. Surface water behind the wall should not be allowed to discharge directly into the wall backfill materials. In addition, water should not be allowed to discharge or pond around walls. Weep holes should be incorporated into the cast-in-place retaining walls to provide for drainage of backfill should groundwater elevations rise. A geosynthetic filter fabric should be placed behind weep holes to reduce the potential for soil loss.

Surcharge loads such as vehicular traffic and construction equipment will induce lateral loads on retaining walls and buried structures. Surcharge-induced lateral pressures acting on retaining structures should be computed using the methods presented in AASHTO using earth pressure design parameters provided in the tables below.

TABLE 5

RECOMMENDED DESIGN PARAMETERS FOR ROCKERY WALLS

Design Parameter Recommended Value

(Cut and Fill)

Total Unit Weight of Backfill or Native 120

Soil Friction Angle, (degrees) 34

Soil Cohesion, c (psf) 0 Wall Backfill Interface

Friction Angle, 23

TABLE 6

RECOMMENDED DESIGN PARAMETERS

FOR CAST-IN-PLACE WALLS

Design Parameter Recommended Value

(Cut and Fill) Total Unit Weight of Backfill or Native 120

Soil Friction Angle, (degrees) 34 Soil Cohesion, c (psf) 0

(degrees) 0

6.3.2 Bearing Resistance

The rockery and cast-in-place retaining wall foundations are considered to be continuous strip foundations with a large length to width ratio. The bearing resistance, including resistance factors, is provided in Figure 10. For our bearing resistance analysis, we used elastic deformation methods to calculate the load at which an allowable settlement is reached. For the service limit bearing resistance that corresponds to 1 inch of static settlement the subgrade modulus is 50 pounds per cubic inch.

In order to provide uniform support for the rockery walls, we recommend proof-rolling and recompacting the subgrade and then placing a minimum of 6 inches of free-draining granular Structural Fill on the subgrade beneath the walls before placing the base rock layer. This layer should extend laterally at least 6 inches in front of the wall and 12 inches behind the wall.

6.3.3 Sliding Resistance

The resistance of the rockery and cast-in-place walls to sliding is provided by friction along the base of the wall. The coefficient of friction between the base of the wall and the underlying Structural Fill is provided in Table 7. We recommend ignoring passive resistance. In the design of the rockery walls where passive resistance is ignored, a sliding resistance factor of

1.0 should be applied for static conditions and 1.0 for seismic (AASHTO). For the design of cast-in-place walls, a sliding resistance factor of 0.8 should be applied for the strength limit state and 1.0 for the extreme limit state (AASHTO).

TABLE 7

RECOMMENDED PARAMETERS FOR

ROCKERY AND CAST-IN-PLACE SLIDING

Parameter Value

Unit weight of soil, soil (pcf) 120 Unit weight of rockery, rock (pcf) 150

Soil friction angle, (degrees) 34 Friction angle for soil to Structural Backfill (degrees) 34

Friction angle for concrete cast against Structural Backfill (degrees) 34 Interlocking Friction Angle for Rock-Rock (degrees) 35

6.3.4 Wall Drainage

Based on conditions encountered in the subsurface explorations, we do not anticipate the presence of groundwater near the base of the proposed rockery walls. Nevertheless, surface water can make its way into the backfill, regardless of the permeability of the material.

Therefore, we recommend providing drainage measures that reduce the potential for water to accumulate in the backfill and for hydrostatic pressures to act on the wall. The final drainage measures should collect water from the base of the wall and not allow water to collect in a bathtub-type situation.

We recommend sloping the ground surface in front of walls a minimum of 5 percent away from the wall face for a minimum horizontal distance of 10 feet measured from the face of the wall (or until a paved surface is encountered, whichever is less). Also the portion of the excavation in front of the base rocks should be backfilled with Structural Fill.

6.4 Pavements

Pavement recommendations for the project are provided in the Fanny Bridge report.

7.0 CONSTRUCTION AND MATERIALS SPECIFICATIONS

The applicability of the design parameters in Section 6.0 is contingent on good construction practice. Poor construction techniques may alter conditions from those upon which our recommendations are based, and therefore result in poor performance. Our analyses assumed that this project is constructed according to FP-14. The following sections provide additional construction considerations for this project.

7.1 Site Preparation

Subgrade for supporting the approach embankments and spread footings should be cleared of trees, shrubs, and any other vegetation; grubbed of stumps and large roots; and stripped of any topsoil and underlying soils that contain significant amounts of roots or other organic matter in accordance with Section 201 and 203 of FP-14. Any rubble or debris should be removed from the subgrade. This subgrade preparation should extend at least 3 feet beyond the toe of the new roadway embankments.

After stripping off topsoil, we recommend preparing the footing and embankment subgrade as follows:

The bottom of bridge spread footing foundations should extend a minimum of 2 feet below lowest adjacent finished grade for frost protection. The subgrade removal and recompaction described below should extend laterally beyond the edge of the footing a distance equal to the footing excavation depth.

Proof-roll the footing and embankment subgrade soil with a heavily loaded dump truck or vibratory roller (or equivalent).

Remove any soft/loose zones noted during rolling and replace those areas with structural fill material constructed in accordance with the recommendations presented below.

Compact the upper 2 feet of the exposed subgrade to at least 95 percent of its maximum dry density determined using AASHTO T-99 Method C.

Place a minimum of 6 inches of free-draining granular Structural Fill on the subgrade beneath the abutment footings

7.2 Backfill Material, Placement, and Compaction

Prior to the placement of any fill or roadway construction, any standing water should be drained from the area and the exposed subgrade proof-rolled as described in Section 7.1. Foundation fill should be placed on the proof-rolled subgrade material. We recommend using the foundation fill as specified in Section 704.01 of FP-14.

Foundation fill should not contain organics or deleterious material. It should be placed in horizontal lifts and compacted to at least 95 percent of its maximum dry density determined using AASHTO T-99 Method C and the thickness of compacted lifts should not exceed 6 inches.

Trench backfill should be placed in lifts not exceeding 4 inches if compacted with hand-operated equipment, or 8 inches if compacted with heavy equipment. Each lift should be compacted to at least 92 percent of the maximum dry density (AASHTO T-99 Method C) 12 inches or more below the pavement subgrade, and to 95 percent within 12 inches of the pavement subgrade. We recommend a minimum cover over utilities of 2 feet from the crown of the pipes or conduits to the top of the pavement subgrade. Catch basins, utility vaults, and other structures installed flush with the pavement should be designed and constructed to transfer wheel loads to the base of the structure.

7.3 Temporary Groundwater Control

Throughout the proposed excavation areas, the on-site soil contains enough fine-grained material to make it moisture sensitive, and groundwater could be encountered in the excavations;

therefore, control of surface and groundwater would be necessary to maintain the integrity of the foundation subgrade material and a firm working platform.

Construction for the abutment footing key may encounter groundwater. The elevation of groundwater at the site is variable and depends on the flows released by the Bureau of Reclamation. The Contractor should be responsible for the control of groundwater and surface water within the contract limits. In this regard, sloping, slope protection, ditching, sumps, dewatering, and other measures should direct water away from work areas.

7.4 General Excavation, Temporary Cut and Permanent Slopes

To provide safe working conditions and prevent ground loss, excavation slopes should be the responsibility of the Contractor, because the Contractor would be at the job site to observe and control the work. All current and applicable safety regulations regarding excavation slopes and shoring should be followed.

Excavations could be accomplished with conventional excavating equipment, such as a dozer, front-end loader, excavator, or backhoe. For planning purposes, we recommend that temporary, unsupported, open-cut slopes be no steeper than 2 Horizontal to 1 Vertical (1H:1V) in the near-surface soil. Flatter cut slopes could be required where looser soil or seepage zones are encountered. We recommend that all exposed cut slopes be protected with a waterproof covering during periods of wet weather to reduce sloughing and erosion. We recommend that permanent slopes be no steeper than 2H:1V.

Excavated material (or stockpiles of construction materials or equipment) should not be placed closer to the edge of any excavation than the depth of the excavation, unless the excavation is shored and such materials are accounted for as a surcharge load on the shoring system.

7.5 Utilities

Utility trenching should be accomplished in accordance with FP-14. We anticipate that excavation could be accomplished with conventional excavation equipment. If debris is encountered during trench excavation, it may be necessary to increase the thickness of bedding material below utilities to maintain a sufficient thickness above the debris in some areas. Soil exposed at the bottom of the deep trenches may be easily softened or disturbed by construction equipment and operations, especially near the groundwater table. The Contractor should take all necessary steps to protect the subgrade from becoming disturbed. If the subgrade is disturbed due to soft or wet conditions, additional soil excavation below the bedding level is recommended. If the subgrade is relatively dry, the excavated soil can be replaced with additional foundation stabilization material. If wet conditions are present in soft compressible soil after over-excavation, then we recommend placing 6- to 12-inch quarry spalls to stabilize the subgrade, before placing foundation stabilization material.

Backfilling of utility trenches should be completed using structural fill compacted as specified above. There should be sufficient cover over the pipe, however, so that when heavy compactors are used, the pipe is not damaged during backfill compaction.

8.0 PLAN REVIEW AND CONSTRUCTION OBSERVATION

Geotechnical design recommendations are developed from a limited number of explorations and tests. Therefore, recommendations may need to be adjusted in the field. To this end, we recommend that a construction observation and monitoring program be implemented for the project and that Shannon & Wilson be retained to observe the geotechnical aspects of construction. This observation would allow us to confirm that conditions encountered are consistent with those used in our analyses. At the same time, it will allow us to provide expedient recommendations should conditions be revealed during construction that are different from those anticipated.

9.0 LIMITATIONS

This report has been prepared for the exclusive use of CH2M Hill and CFLHD for the purpose of providing geotechnical recommendations for the referenced project. It should be made available to prospective contractors for the purpose of providing information on factual data only, and not as a warranty of subsurface conditions, such as those interpreted from the logs and discussions of subsurface conditions in this report.

The analyses, conclusions, and recommendations contained in this report are based on site conditions as they existed at the time of our explorations. If there is a substantial lapse of time between the submission of this report and the start of construction at the site, if alternative design codes are adopted, or if conditions change due to natural causes or construction operations at or adjacent to the site or appear different from those described in our report, we recommend that we be retained to review our report to determine the applicability of the conclusions and recommendations considering the changed conditions and time lapse.

Caltrans provided review comments on the geotechnical reports submitted as the project progressed. Comments on the report as submitted by Caltrans including our responses are included in Appendix E.

Within the limitations of scope, schedule and budget, the analyses, conclusions and recommendations presented in this report were prepared in accordance with generally accepted professional geotechnical and geological principles and practice in this area at the time this report was prepared. We make no other warranty, either express or implied.

Unanticipated soil conditions are commonly encountered and cannot be fully determined by a limited boring and testing program. Such unexpected conditions frequently require that additional expenditures be made to attain a properly constructed project. Therefore, some contingency fund is recommended to accommodate such potential extra costs.

The scope of our services did not include an evaluation regarding the presence or absence of hazardous or toxic materials in the soil, surface water, groundwater, or air, on or below or around this site. If such contamination exists, it would not be possible to document it within this limited scope of work.

Shannon & Wilson, Inc. has prepared Appendix F, “Important Information About Your Geotechnical Report,” to assist you and others in understanding the use and limitations of our reports.

SHANNON & WILSON, INC.

Robert A. Mitchell, P.E., G.E.

Vice President

KPC:RAM/kpc

05/20/16

10.0 REFERENCES

American Association of State Highway and Transportation Officials (AASHTO), 2012, AASHTO LRFD bridge design specifications: customary U.S. units (6th ed.): Washington, D.C., AASHTO, 2 v.

Caltrans, Bridge Design Specifications, accessed online January 8, 2016:

http://www.dot.ca.gov/hq/esc/techpubs/manual/bridgemanuals/bridge-design-specifications/bds.html.

Caltrans, Geotechnical Manual, accessed online January 8, 2016:

http://www.dot.ca.gov/hq/esc/geotech/geo_manual/manual.html.

Caltrans, ARS Online, Response Spectra Calculator, Version 2.3.06, 2013, (http://dap3.dot.ca.gov/ARS_Online/index.php).

Central Federal Lands Highway Division (CFLHD), 2010, CFLHD Supplement 11.4.2-1, Aggregate Surfacing Design Standards and Guidance, November, available from http://www.cflhd.gov/resources/design/supplement/ch11/.

Central Federal Lands Highway Division (CFLHD), 2006, Rockery Design and Construction Guidelines, Publication No. FHWA-CFL/TD-06-006, November.

U.S. Department of Transportation (USDOT) and Federal Highway Administration (FHWA), 2008, Federal Lands Highway Project Development and Design Manual: U.S. Department of Transportation and Federal Highway Administration, available from http://flh.fhwa.dot.gov/resources/manuals/pddm/.

U.S. DOT and FHWA, 2014, Federal Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects, FP-14, U.S. Department of Transportation and Federal Highway Administration, available from http://flh.fhwa.dot.gov/resources/specs/.

Wood Rodgers, 2011, Preliminary Geotechnical Report, Proposed Fanny Bridge, #19-0033, Tahoe…