160603-Fanny_Bridge_and_Pavement_Geotechnical_Report-final.pdf

Geotechnical Report Central Federal Lands Highway Division

CA SR 89/SR 28 Intersection Area Fanny Bridge Replacement

Bridge No. 19-0033, Placer County, Route 89, PM 8.48 Caltrans EA: 03-3A7601

Caltrans PIN: 0300000452 Tahoe City, California

May 20, 2016

Submitted To:

Mr. Bill Lang

CH2M Hill, Inc.

9191 South Jamaica Street

Englewood, CO 80112

By:

Shannon & Wilson, Inc.

400 N 34th Street, Suite 100 Seattle, Washington 98103

23-1-01373-205 i

TABLE OF CONTENTS

Page

1.0 INTRODUCTION

2.0 PROJECT AND SITE DESCRIPTION

3.0 FIELD EXPLORATIONS AND LABORATORY TEST RESULTS

3.1 Subsurface Explorations

3.2 Laboratory Test Results

4.0 SUBSURFACE CONDITIONS

4.1 Pavement Borings B-1-15 to B-6-15 and B-9-15 to B-11-15

4.2 Bike Path and Trail Retaining Wall Boring B-2-15

4.3 Fanny Bridge Borings B-7-15 and B-8-15

4.4 Groundwater

4.5 Scour

5.0 GEOLOGY AND GEOLOGIC HAZARDS

5.1 Regional Geology

5.2 Geologic Hazards

5.2.1 Seismic Ground Motion

5.2.2 Liquefaction

5.2.3 Fault Hazard

6.0 CONCLUSIONS AND RECOMMENDATIONS

6.1 Abutment Pile Foundations

6.1.1 Axial Resistance

6.1.2 Estimated Pile Settlements

6.1.3 Lateral Resistance

6.2 Pavement Design

6.2.1 General

6.2.2 Traffic Loading

6.2.3 Subgrade Soils

6.2.4 Full Depth Reclamation

6.2.5 Pavement Sections

6.3 Abutment Wall Design

6.4 Other Retaining Wall Design

TABLE OF CONTENTS (cont.)

ii

7.0 CONSTRUCTION AND MATERIALS SPECIFICATION

7.1 Driven Pile Installation

7.2 Site Preparation

7.3 Backfill Material, Placement, and Compaction

7.4 Temporary Groundwater Control

7.5 General Excavation, Temporary Cut and Permanent Slopes

7.6 Utilities

8.0 CONSTRUCTION OBSERVATION

9.0 LIMITATIONS

10.0 REFERENCES

TABLES

1 Scour Data 2 Design Response Spectrum 3 Pile Data 4 Fanny River Bridge FB Multi-Pier Soil Input Parameters 5 Traffic Loading 6 Recommended Pavement Sections 7 Recommended Soil Parameters for Abutment Wall Design

FIGURES

1 Vicinity Map 2 Site and Exploration Overview 3 Truckee River Bridge Site and Exploration Plan 4 Fanny Bridge Site and Exploration Plan 5 Design Response Spectrum 6 Liquefaction Analysis Summary Boring B-7-15 M=6.8, PGA=0.43 7 Liquefaction Analysis Summary Boring B-8-15 M=6.8, PGA=0.43 8 Fanny River Bridge Axial Resistance 24 inch dia. Closed End CISS Pile

APPENDICES

A Subsurface Explorations B Laboratory Test Results C Response to Caltrans Comments D Important Information About Your Geotechnical Report

GEOTECHNICAL REPORT

CENTRAL FEDERAL LANDS HIGHWAY DIVISION

CA SR 89/SR 28 INTERSECTION AREA

FANNY BRIDGE REPLACEMENT

BRIDGE NO. 19-0033, PLACER COUNTY, ROUTE 89, PM 8.48

CALTRANS EA: 03-3A7601, CALTRANS PIN: 0300000452

TAHOE CITY, CALIFORNIA

1.0 INTRODUCTION

This geotechnical report provides geotechnical recommendations towards delivery of a final set of plans, specifications and estimate for the Federal Highway Administration, Central Federal Lands Highway Division (CFLHD) California State Route (SR) 89/ SR 28 intersection area near Tahoe City, California (Figure 1). Our work was completed in general accordance with our scope of work and purchase order agreement with CH2M Hill, Inc. (CH2M Hill) dated August 17, 2015 and modified September 30, 2015.

This report presents recommendations for the design and construction of the Fanny Bridge replacement and includes subsurface explorations, laboratory testing, and design and construction recommendations. Additionally, this report includes pavement recommendations for the CA SR 89/SR 28 Intersection project. The recommendations for the design and construction of the Truckee River Bridge are presented in a separate report.

2.0 PROJECT AND SITE DESCRIPTION

The SR 89 route is currently the only transportation route along Lake Tahoe’s west shore for visitors making connections from the national highway system, US 50 and Interstate 80. The two lanes SR 89 along the west shore also provides the only access to over 25,000 acres of federal lands managed by the United States Forest Service (USFS) – Lake Tahoe Basin Management Unit. The historic Fanny Bridge is narrow at 2 lanes and does not provide adequate access for the 22,000 vehicles per day and 400 bike and pedestrians per hour during peak summer months with visitors accessing the USFS and State Parks recreational areas. Backups in Tahoe City at Fanny Bridge have extended over 2 miles south down SR 89 equating to 2 hour or more delays. Congestion issues also pose safety risks in emergency response time delays. The proposed improvements would enhance the Tahoe City and west shore community with improved access to federal lands and enhancing bike and pedestrian facilities removing bike/pedestrian traffic from the narrow SR 89 shoulders.

Project features as shown in Figure 2 include:

A new bridge over the Truckee River and associated approaches Replacement of the existing Fanny bridge New bike baths under the new Truckee River bridge that to tie into existing trails Construction of east, west, and WYE roundabouts Pavements associated with the proposed bridges and roundabouts Retaining walls (i.e. rockery, Type 1 cast-in-place, etc.)

3.0 FIELD EXPLORATIONS AND LABORATORY TEST RESULTS

3.1 Subsurface Explorations

Between October 12 and October 15, 2015, Taber Drilling, under subcontract to Shannon & Wilson, drilled 11 borings to depths ranging between 4.5 and 101.5 feet. Borings were drilled using hollow-stem augur and mud-rotary drilling techniques with Standard Penetration Tests and Modified California samples collected at 2.5 to 5-foot intervals. The borings for the Fanny Bridge, B-7-15 and B-8-15, and the retaining wall boring, B-2-15, were drilled with mud rotary drilling techniques and the remaining were drilled using a hollow-stem auger. A representative from Shannon & Wilson was present throughout the field explorations to observe drilling operations, retrieve representative soil samples for laboratory testing, and prepare descriptive field logs of the explorations. Per AASHTO specifications, the hammer energy is not required to be measured and may be assumed to be 80 percent for the auto-hammer utilized during drilling.

The locations of the borings are shown in Figures 2 through 4 along with locations of previous explorations and ReMi geophysical by others. Appendix A presents the logs of the borings drilled for this study.

3.2 Laboratory Test Results

Laboratory tests performed on samples from the borings consist of moisture content, grain-size distributions, Atterberg limits, direct shear tests, and soil corrosivity tests. The laboratory test results are presented in Appendix B.

4.0 SUBSURFACE CONDITIONS

The general subsurface conditions observed during Shannon & Wilson explorations are presented in the following sections. The purpose of the borings drilled for this scope of services was to evaluate the surface conditions for the project-wide pavement design, a retaining wall that runs parallel to the bike path and trail near the Caltrans Maintenance Station (Figure 2), and foundation support for the Fanny Bridge.

4.1 Pavement Borings B-1-15 to B-6-15 and B-9-15 to B-11-15

Borings B-1-15 to B-6-15 and B-9-15 to B-11-15 were drilled to access the project-wide pavement conditions. The asphalt concrete encountered in these borings ranged between 12 and 26 inches thick. Along the proposed paved roadway in a majority of the borings the surficial soils consisted of up to 6 1/2 feet of medium dense, silty sand with some gravel layers. We observed medium stiff to very stiff sandy lean clay to lean clay underlying the silty sand in borings B-4-15 through B-6-15. We did not observe groundwater in any of these pavement borings.

4.2 Bike Path and Trail Retaining Wall Boring B-2-15

Near the proposed retaining wall for the bike bath and trail, the generalized subsurface profile consists of the following soil layers:

A fill layer that is 10 1/2 feet thick containing medium dense, clayey sand with gravel and stiff to hard, sandy fat clay

Medium dense to very dense, silty sand layer that extends to 25 1/2 feet below ground surface

Groundwater was observed at approximately 11 feet during drilling

4.3 Fanny Bridge Borings B-7-15 and B-8-15

Borings B-7-15 and B-8-15 were drilled at the abutments of Fanny Bridge and the idealized subsurface profile consists of the following soil layers:

A fill layer that is 4 1/2 to 8 1/2 feet thick containing medium dense, silty sand with gravel

Underlain by 2 1/2 to 3 1/2 feet of medium stiff to very stiff, lean clay

Underlain by 6 to 11 feet of medium dense to dense, poorly graded gravel consisting of fine gravel

Alternating layers of stiff to hard, fat clay and very dense poorly graded gravel to the remaining depth of the explorations

Groundwater was observed during drilling at a depth between 7 and 15 feet

4.4 Groundwater

The groundwater level was observed in the Fanny Bridge explorations B-7-15 and B-8-15 at a depth of 7 to 15 feet on both sides of the river. For evaluation purposes, the groundwater elevation was assumed to be equal to the Truckee River which is about 6,220 feet. We understand that the river elevation varies depending on the flow allowed by the Bureau of Reclamation.

4.5 Scour

Table 1 presents the scour data for the bridge abutments. The scour data in Table 1 was obtained from the project hydraulic report and project drawings prepared by others.

TABLE 1

SCOUR DATA

100-year

Support No.

Long Term Scour (degradation, ft)

Short-Term Scour (ft) General Contraction Local

Depth Elevation Depth Elevation 1Depth Elevation Abutment 1 0.00 6215.73 0.50 6215.23 4.11 6211.62 Abutment 2 0.00 6215.73 0.50 6215.23 4.11 6211.62

500-year

Support No.

Long Term Scour (degradation, ft)

2Short-Term Scour (ft) General Contraction Local

Depth Elevation Depth Elevation Depth Elevation Abutment 1 0.00 6215.73 4.34 6211.39 3.65 6212.08 Abutment 2 0.00 6215.73 4.34 6211.39 3.65 6212.08

Notes: 1) Per NCHRP methodology for local abutment scour, general contraction scour values are not additive, but instead are already included.

2) For abutments only. Per current NCHRP methodology and coordination with FHWA, local scour depth at abutments includes general horizontal (non-pressure) scour only and is not additive to long-term degradation or general contraction scour. Vertical contraction scour depth due to pressure is greater than the NCHRP local scour depth, and is therefore used as the maximum scour depth (riprap countermeasures per HEC-23 DG 14 are also included).

5.0 GEOLOGY AND GEOLOGIC HAZARDS

5.1 Regional Geology

The project area is located near the western shoreline of Lake Tahoe along the eastern extents of the Sierra Nevada Mountain Range, which comprises a tilted fault block approximately 400 miles long. Mesozoic granite rock forms the Sierra Nevada Batholith, which underlies the Tahoe Basin (Hyne, 1972). The granitic rock along the perimeters of Lake Tahoe is extensively weathered due to extreme water fluctuations from recurrent glacial damming in the Pleistocene.

Tertiary volcanic rocks overly the granites and includes basalt, andesite, latite, trachytes, tuff, and mudflows. These tertiary units are frequently observed at the ground surface (Grose, 1985).

Other deposits that are often observed locally are alluvium, colluvium, and glacial materials.

5.2 Geologic Hazards

5.2.1 Seismic Ground Motion

The seismic analysis is based on ground motion levels, performance criteria, and methods as described in American Association of State Highway and Transportation Officials (AASHTO;

6th Edition, 2012) and the Caltrans Amendments to AASHTO. The site class is determined using standard penetration test (SPT) N-values and/or shear wave velocities measured in the upper 100 feet of the soil profile. The soil classification at the Fanny Bridge location was derived based on explorations B-7-15 and B-8-15 completed near the proposed bridge. The average SPT N-value for the upper 100 feet of borings B-7-15 and B-8-15 is 25 blows per foot. Using the AASHTO site class criteria, the subsurface SPT N-values and shear wave velocities near the Truckee River and Fanny Bridge indicate Site Class D should be used in design.

The design response spectrum corresponds to a 975-year return period ground motion and was determined for the project site using the Caltrans ARS Online application (Version 2.3.06), which considers both deterministic and probabilistic seismic hazard methods. The design response spectrum is based on the larger of deterministic and probabilistic spectral accelerations.

The design response spectrum incorporates Site Class D soil effects using a Vs30 equal to about 900 feet per second. The design response spectrum is presented in Figure 5 and numerical values are presented in Table 2.

TABLE 2

DESIGN RESPONSE SPECTRUM

Period (seconds)

Spectral Acceleration (g)

Period (seconds)

Spectral Acceleration (g)

0.01 0.43 0.70 0.70

0.05 0.66 0.85 0.64

0.10 0.79 1.00 0.59

0.15 0.90 1.20 0.51

0.20 0.99 1.50 0.43

0.25 0.97 2.00 0.32

0.30 0.95 3.00 0.20

0.40 0.86 4.00 0.14

0.50 0.80 5.00 0.11

0.60 0.74 Notes: 1) Spectral accelerations are the larger of the deterministic and probabilistic values obtained from

Caltrans ARS Online.

2) Spectral values include near fault factors as incorporated by Caltrans ARS Online.

3) Spectral values are based on a Vs30 of about 900 ft/s.

5.2.2 Liquefaction

Based on AASHTO and the Caltrans Geotechnical Manual, we estimated the liquefaction potential at the Truckee River Bridge for the 975-year return period ground motion. The liquefaction analysis described below is based on Youd et al. (2001) and provides the consensus of the profession up to the year 2000. More recent liquefaction analysis methods are available and could provide a slightly different estimate of liquefaction potential; however, the Youd et al.

(2001) method is typically considered to be the benchmark method for other liquefaction analysis methods and is recommended by AASHTO and the Caltrans Geotechnical Manual (2014).

For liquefaction potential of low plasticity silt we considered the methods recommended by Idriss and Boulanger (2008) and Bray and Sancio (2006). The Caltrans Geotechnical Manual (2014) recommends the Bray and Sancio (2006) method and both methods are included in

AASHTO.

The peak ground surface acceleration for this site is 0.43g. Based on the USGS seismic hazard deaggregation for the site the magnitude appropriate for the 975-year return period ground motion and corresponding liquefaction analysis is about 6.8. This magnitude is associated with the North and West Tahoe faults that are about 3 to 6 miles (5 to 9 kilometers) from the site.

Liquefaction of loose, saturated, cohesionless soils due to seismic loading has been studied extensively over the past 40 years, resulting in both laboratory and field procedures to evaluate liquefaction potential. The analyses presented in this report are derived in part from the recommendations presented in Seed and Idriss (1971), NCEER (1997), Youd et al (2001), Robertson and Wride (1998), and Robertson (2009). Liquefaction potential is reported in terms of a factor of safety (FS) representing the ratio of soil resistance to seismic demand. Soil that has a FS against liquefaction greater than 1 is suitable for the 975-year return period ground motion.

DemandSeismic ResistanceSoilFS

The soil’s resistance to liquefaction is termed cyclic resistance ratio (CRR). For the site specific SPT data, the CRR is estimated using the empirical relationship by Youd et al (2001).

This empirical relationship relates N1,60,CS to CRR for a moment magnitude of 7.5 at a vertical effective stress of 1 atmosphere (atm), or CRR7.5, 1atm. The field blow count values, N, were corrected for energy, overburden stress, and fines content per the procedures in Youd et al (2001).

Hammer energy for the auto-hammer used during drilling was assumed to be 80 percent according to AASHTO. The cone penetration test data was interpreted utilizing the procedures described in Robertson (2009).

Since the magnitude associated with the 975-year return period earthquake is not equal to 7.5, the CRR7.5 was adjusted by means of a magnitude scaling factor (MSF) as follows:

where the MSF recommended in Transportation Research Board (TRB; 1999), also referred to as the “lower bound” in Youd et al, was chosen for this project and is calculated as:

= 10 ..

The seismic demand is termed cyclic stress ratio (CSR) and is expressed as the ratio of average equivalent shear stress to the initial vertical effective overburden stress. The cyclic stress ratio of soil at depth in a soil deposit is estimated by the following equation (Seed and Idriss, 1971):

d vo vo vo r g aCSR ' max

' 65.0 where amax is the peak ground acceleration generated at the ground surface, g is the acceleration of gravity, vo and 'vo are the total and effective vertical overburden stresses, respectively, and rd is a stress reduction coefficient. The stress reduction coefficient adjusts the ground surface PGA for the response of the soil profile as a function of depth. The relationship of rd as reported by Liao and Whitman (1986) and Youd et al (2001), which represents the average values reported by Seed and Idriss (1971), was used to calculate the stress reduction coefficient.

The liquefaction potential of the soil encountered in borings B-7-15 and B-8-15 was evaluated. The results of this evaluation indicate that the site soil would not liquefy during the design ground motion. The liquefaction potential of the low plasticity sandy silt and silty sand includes consideration of the behavior of low plasticity silt based on the Idriss and Boulanger (2008) criteria. Figures 6 and 7 provide plots of factor of safety versus depth of the liquefaction analysis performed for borings B-7-15 and B-8-15.

Using the results of the liquefaction analysis presented above, we estimate that post-seismic abutment and approach settlement would be about relatively negligible for borings B-7- 15 and B-8-15.

5.2.3 Fault Hazard

There are three nearby faults that are classified as Quaternary Class A or Class B faults.

The Incline Village fault is less than 15,000 years old and is located approximately 9 miles to the east of the site. The North Tahoe Fault is less than 15,000 years old and is located approximately 5 miles east of the site. The West Tahoe-Dollar Point fault is less than 130,000 years old and is located approximately 3 miles from the site. The 1972 Alquist-Priolo Earthquake Zoning Fault Act has not designated the locale as an Earthquake Fault Zone (EFZ).

6.0 CONCLUSIONS AND RECOMMENDATIONS

The proposed Fanny Bridge will replace the existing bridge with a 128-foot long and 44-feet wide single span bridge and the proposed abutments will be located behind the existing bridge abutments. Abutments 1 and 2 are on the south and north sides of the river, respectively. The bridge deck elevation is approximately 6,233 and 6232 feet at the south and north abutments, respectively. The 100- and 500-year high water elevations are 6,226 and 6,228 feet, respectively.

6.1 Abutment Pile Foundations

Based on the subsurface conditions encountered in the field explorations and our understanding of the project, engineering studies were performed to develop recommendations for design of the proposed structure foundations. The Fanny Bridge project site is underlain by medium dense silty sand and gravel, stiff clay, and very dense clayey gravel and gravelly clay. These deposits extend to the full depth of the two explorations completed for the bridge. Due to the presence of a boulder or cobble, boring B-7-15 on the north side of the bridge was terminated at a depth of 57 feet. However, on the south side of the bridge, boring B-8-15 extended to a depth of 100 feet.

Deep foundations that extend into the underlying very dense clayey gravel (north) and gravel (south) are recommended to support the proposed replacement bridge.

We understand that cast-in-steel-shell (CISS) piles are the preferred deep foundation for this project. The proposed CISS piles consist of driving a closed end steel casing and filling this casing with reinforced concrete. Geotechnical engineering recommendations for 24-inch-diameter CISS piles are provided below.

6.1.1 Axial Resistance

Axial resistance analyses were performed for the 24-inch-diameter CISS piles. The service, strength, and extreme limit state resistance for the 24-inch-diameter CISS piles were determined based on soil types encountered in the borings and relative densities of these soils as determined by Standard Penetration Test (N-values). The axial resistance is based on AASHTO (6th Edition, 2012) and the Caltrans Amendments to AASHTO and is a sum of the side resistance and end bearing at the pile tip. Results of our axial resistance analyses are presented graphically in Figure 8 as a plot of pile penetration versus axial compressive capacities for the service and strength limit state conditions. The recommended resistance factors for the service, strength, and extreme limit state that would be applied to the nominal resistance are provided in Figure 8.

Generalized subsurface conditions for the borings are also presented in the figure. The recommended preferred pile size and recommended lengths are presented in Table 3.

The nominal bearing resistance of a pile group bearing in the cohesionless soil encountered at the site is the sum of the piles in the group for center-to-center pile spacing of 2.5 diameters or greater (AASHTO 10.7.3.9)

TABLE 3

PILE DATA

Location Pile Type

Nominal Resistance (kips) Design Tip Elevation

(ft)

Specified Tip

Elevation (ft)

Required Nominal Driving

Resistance (kips)Compression Tension

Abut 1 24-inch Dia. CISS (24x0.375) 500 200

6183 (a) n/a (b)

6183 (c) 6192 (d)

6183 500

Abut 2 24-inch Dia. CISS (24x0.375) 500 200

6182 (a) n/a (b)

6182 (c) 6191 (d)

6182 500

Notes: 1) Design tip elevations are controlled by: (a) Compression, (b) Tension, (c) Settlement, (d) Lateral Load.

2) A WEAP analysis and pile driving criteria should be performed after the pile hammer is selected by the Contractor.

The specified tip elevation assumes that dynamic testing is performed on at least two piles per site condition and no less than 2% of the production piles. In our opinion, at least one pile should be dynamically tested using PDA/CAPWAP at each Fanny River Bridge abutment for the one site condition observed at the Fanny Bridge.

6.1.2 Estimated Pile Settlements

The pile settlement analysis is based on the subsurface conditions at the site, axial loads, and the assumption that the piles would extend into dense to very dense gravelly soil at a depth of about 50 to 55 feet below the existing ground surface. Settlement of piles under static loading conditions would typically be due to the elastic compression of the pile, elastic settlement of granular soils, and consolidation of cohesive soils located along the pile shaft and below the pile tips. Settlements of the pile group as a result of service loading is estimated based on the results of the field and laboratory testing and the service pile loads. Modulus values are estimated for the granular and silt soils using the shear wave velocity estimates and SPT N-values collected in the soil borings. A simplified 1H:2V stress distribution beginning at about two-thirds of the depth of the pile group penetration was considered using the service pile loads. Based on this analysis the Fanny Bridge pile group could settle between 1 and 1 ½ inches during the service load condition.

6.1.3 Lateral Resistance

Lateral loads acting on the structure from wind, seismic events, and other loadings may be resisted by the lateral resistance provided by the piles. The computer program FB MultiPier will be used to generate lateral p-y curves for the lateral resistance analysis of piles and to calculate the magnitude of deflection, shear, and moment along the pile. Based on subsurface conditions as interpreted from the explorations accomplished at the bridge, the recommended parameters for input into the FB MultiPier under static and seismic loading conditions are presented in Table 4.

The recommended efficiency factors or pile load modifiers for the p-y curves due to pile group action are listed in both AASHTO and Caltrans Amendments to AASHTO Table 10.7.2.4- 1.

6.2 Pavement Design

6.2.1 General

The recommended pavement sections were calculated using the methodology described in the California Department of Transportation Highway Design Manual (HDM; Caltrans, 2015).

This methodology takes into consideration the anticipated traffic loading, type of subgrade soils, and pavement section material properties when calculating a design section.

SHANNON &WILSON, INC.

Depth below ground surface to top of the soil layer

Depth below ground surface to bottom of the soil layer

Ground Water Depth1

Soil Type Total Unit Weight

Poisson's Ratio

Young's Modulus

Shear Modulus

Peak Friction Angle

Undrained Shear

Strength Lateral Model Axial Model

Torsional Model

Tip Model

Averaged Undrained

Shear Strength

Strain at 50% Max

Steel Pile Ultimate Unit Skin Friction4

Torsional Shear

Stress 2

Steel Pile Axial

Bearing Capacity3,4

(feet) (feet) (feet) (pcf) (ksi) (ksi) (degrees) (psf) Static Cyclic (psf) (psf) (psf) (ksf)

SC/GC/GP 0.0 20.0 Cohesionless 115 0.33 5.3 2.0 33 - Sand (Reese) Driven Pile Hyperbolic Driven Pile 50 50 - - 260 260 36

CL 20.0 45.0 Cohesive 110 0.42 14.1 5.0 - 2,200 Clay (Soft < Water) Driven Pile Hyperbolic Driven Pile - - 2,200 0.005 1,100 1,100 20

GP/GC 45.0 60.0 Cohesionless 125 0.33 35.9 13.5 40 - Sand (Reese) Driven Pile Hyperbolic Driven Pile 125 125 - - 1,100 1,100 96

CL 60.0 - Cohesive 110 0.42 21.3 7.5 - 2,600 Clay (Soft < Water) Driven Pile Hyperbolic Driven Pile - - 2,600 0.005 1,300 1,300 24

Notes:

1. Depth of water based on ground surface at elevation at borings B-7-15 and B-8-15.

2. Torsional shear stress is assumed to be equal to the unit side resistance.

3. Program requires the axial bearing failure in kips, need to multiply the bearing capacity by the cross sectional area at the tip of the pile.

4. Given parameters assume a driven steel pile and are not applicable to alternative types of deep foundations.

7.0

TABLE

Fanny River Bridge FB Multi-Pier Soil Input Parameters

Soil Description

General Parameters Strength Parameters Pile Design Parameters

Modulus of Subgrade Reaction, k

(pci)

For our analysis, we separated the project roads into four segments.

A-Line South – SR 89 South of the Truckee River and North of the New SR89 A-Line North – SR 89 South of the Truckee River and South of the New SR89 B-Line – New SR 89 C-Line – SR 89 and SR 28 North of the Truckee River

In addition to the intersection reconfiguration, we understand that the access road for the Lumber Yard and the Caltrans Maintenance Yard will be reconstructed as part of this project.

6.2.2 Traffic Loading

To estimate a 20 year pavement design life, we used the data presented in the site traffic study report (CH2M Hill/Wood Rogers, 2014) to determine the 18-kip Equivalent Single-Axle Loading (ESAL) values. CH2M’s traffic study estimated the reconfigured intersection annual average daily traffic (AADT) values and indicates approximately 7 percent of the total traffic was composed of trucks. For the distribution of various truck types (i.e., 2-, 3-, 4-, and 5-axle trucks), we were provided the Caltrans measured vehicle type distributions for the SR 89 and SR28. The following truck-type distribution used for each segment is listed below:

2-axle trucks = 62.2% 3-axle trucks = 17.9% 4-axle trucks = 11.5% 5-axle trucks = 8.4%

For the Lumber Yard and Caltrans Maintenance Access Road, the traffic loading is unknown. Based on discussions with CH2M, an annual average daily truck traffic (AADTT) of 100 trucks per day was assumed for this access road. The resulting two-way total ESALs was reduced by applying a directional coefficient of 0.6 and converted to a traffic index (TI). Table 5 below summarized the traffic information for each roadway segment and the resulting TI for each segment.

6.2.3 Subgrade Soils

To evaluate existing subgrade strengths, Hveem stabilometer (R-value) tests were completed on three bulk samples of representative subgrade materials in accordance with ASTM D-2844. The bulk sample tested on soils encountered in borings B-5-15, B-9-15, and B-11-15 resulted in R-values of 17, 79, and 82 respectively. In general, the fill materials encountered north of the Truckee River were relatively sandy clay soils and our borings south of the Truckee River generally consisted of relatively silty sands. For our design we used an R-Value of 17 for

Segment 1 and, due to the potential variability across the alignment, for segments Segment 2 through 4 we selected a design R-Value of 50 for our analysis.

TABLE 5

TRAFFIC LOADING

Segment Roadway

Two-Way

AADT

Two-Way

AADTT

ESAL TI

A-Line North

SR 89 South of Truckee, North of New SR 89 5,400 378 761,000 8.5

A-Line South

SR 89 South of Truckee, South of New SR 89 13,500 945 1,789,000 9.5

B-Line New SR 89 8,800 616 1,240,000 9.0

C-Line SR 89 North of Truckee River and SR 28 12,700 889 1,789,000 9.5

Lumber Yard/

Access Road

Lumber Yard/Caltrans Maintenance Access Road. - 100 201,000 7.5

6.2.4 Full Depth Reclamation

We understand that a pulverized full depth reclamation (FDR) is being considered for rehabilitation of the existing roadway pavement sections. The process involves pulverizing and mixing the existing hot asphalt concrete pavement with the underlying layers to create a uniformly blended, homogeneous material that will be incorporated into the proposed pavement section as a base course. The FDR material is then recompacted, and new asphalt pavement is placed over the FDR. The benefit of the FDR process is that it allows the existing material to be recycled in place and eliminates the expense to export the material offsite. FDR sections typically result in a net overall grade raise from the exiting roadway.

Observed asphalt thicknesses in our boring logs are based on borings drilled in the existing roadway shoulders, and these thicknesses vary from 1 to 2 feet. Typically the FDR process is limited to 8 to 10 inches of existing asphalt due to limitations of both the compaction and the pulverization equipment. In areas where the existing asphalt thicknesses are greater than 10 inches, milling the existing asphalt section can be performed prior to the FDR process. In addition, by milling the existing asphalt, it is potentially feasible to maintain existing grades.

We recommend additional explorations be performed to further characterize existing asphalt thicknesses in drive-lanes. In our opinion, these explorations can be accomplished by the Contractor during construction to verify that the existing roadway pavement section is of sufficient thickness for the FDR process. Final roadway grades may need to be raised if the combined existing base and asphalt thicknesses are less than the proposed asphalt and FDR thicknesses.

6.2.5 Pavement Sections

Based on Caltrans pavement design procedures, we recommend the following pavement sections for the proposed project:

TABLE 6

RECOMMENDED PAVEMENT SECTONS

Segment Pavement Sections Removal and Replacement Full Depth Reclamation

A-Line North 5 in. AC over 6 in. AB 5 in. AC over 8 in. FDR

A-Line South 6-½ in. AC over 6 in. AB 5-½ in. AC over 8 in. FDR

B-Line 6 in. AC over 6 in. AB -

C-Line Alt. 1: 8.0 in. AC over 14 in. AB Alt. 1: 10-½ in. AC over 8 in. FDR

Alt. 2: 6-½ in. AC over 8 in. AB over 8 in. AS Alt. 2: 9-½ in. AC over 10 in. FDR Lumber

Yard Access Road

6-½ in. AC over 10 in. AB -

Note: AC = Asphalt Concrete; AB = Aggregate Base; AS = Aggregate Subbase

We recommend that an asphalt binder of Performance Graded (PG) 64-28 Polymer Modified (PM) be used in the design. The PM grade binder reportedly provides improved resistance to rutting, thermal cracking, fatigue damage, stripping, and temperature susceptibility (Caltrans, 2015). The binder selection is based on the anticipated pavement temperatures, traffic patterns, and traffic speed. A tack coat should be placed between subsequent lifts if the underlying lift will be used for traffic or left uncovered for a significant period of time.

We understand the Contractor will be encouraged to reuse existing site materials to reduce costs associated with importing and exporting new materials. One potential method of generating an aggregate base material is to reuse the millings obtained from the existing asphalt pavement. These millings, referred to as recycled asphalt pavement (RAP), may require field screening to match Caltrans base course gradational specifications. Based on the varying thicknesses of existing asphalt thicknesses observed in the project borings and measured asphalt thickness from utility pot-holing, it is unclear as to the quantity of material that can be generated from the A- and C-Line pavements.

Considering the anticipated adjustments to the roadway profile and the sectional widening of the roadway, FDR was determined to be difficult to construct with the anticipated construction phasing. Additionally, transitioning from the removal and replacement pavement section in the widening areas to the FDR section over a relatively short construction distances was not desirable. The removal and replacement pavement sections will allow for a constant pavement thickness over a given roadway segment, and will eliminate potential complications sectional widening. Based on discussions with CH2M, Caltrans, Placer County, and Central Federal Lands, we understand that the removal and replacement pavement sections are the preferred construction method with a portion of the AB material generated from onsite RAP. For the C-Line, we understand that Alternative 1 (8 inches of AC over 14 inches of AB) will be utilized in the construction plans.

6.3 Abutment Wall Design

Based on the project plans and subsurface conditions from the site investigations, 10 to 15-foot-tall retaining walls are required at the bridge abutment to support the proposed grades.

We recommend that fill material used to backfill the retaining walls consist of free draining, sand and gravel Structural Backfill conforming to Section 704.04 of the 2014 Federal Highways Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects (FP-14). Based on the use of Structure Backfill, the parameters shown in Table 7 are recommended for the design of walls. At rest parameters are included in Table 7 for walls that are not free to deflect. In accordance with the Caltrans Bridge Design Specifications (2004) and the Caltrans Geotechnical Manual (2014), the horizontal coefficient (kh) is equal to one-third peak ground acceleration which is equal to about 0.15g for the project site. The active earth pressure coefficient with earthquake effect is included in Table 7. This value assumes the abutment walls can tolerate displacements of 1 to 2 inches or more during a design seismic ground motion. Based on conversations with the structural engineer, we understand that the abutment walls would be able to displace this amount during the design ground motion.

The parameters provided for the retaining walls assume a free-draining backfill condition. As such, it will be important to control surface water and reduce its potential to infiltrate into the ground around the retaining walls. Surface water behind the wall should not be allowed to discharge directly into the wall backfill materials. In addition, water should not be allowed to discharge or pond around walls. Weep holes should be incorporated into the retaining walls to provide for drainage of backfill should groundwater elevations rise. A geosynthetic filter fabric should be placed behind weep holes to reduce the potential for soil loss.

TABLE 7

RECOMMENDED SOIL PARAMETERS FOR ABUTMENT WALL DESIGN

Design Parameter Recommended Value Total Unit Weight of Backfill (pcf) 120 Friction Angle (degrees) 34 Cohesion 0 Equivalent Fluid Density for Active Retaining Conditions (AASHTO Section 3.11.5.5) (pcf) 34

Equivalent Fluid Density for At-Rest Retaining Conditions (AASHTO Section 3.11.5.5) (pcf) 53

Active Earth Pressure Coefficient with Earthquake Effect (Kae) for Horizontal Backfill (AASHTO Section A11.3.1) 0.37

Note:

a) For the strength limit state, a resistance factor ( ) of 0.8 should be applied to the nominal coefficient of sliding resistance value (AASHTO Sections 11.6.3.6, 10.6.3.4, and Table 10.5.5.2.2-1). For the extreme limit state, a resi coefficient of sliding resistance value.

b) Load factors have not been included in any of the values in this table.

As stated in AASHTO 11.6.5.3, the seismic active earth pressure should be distributed using the same distribution as the static active earth pressure (i.e. triangular distribution increasing with depth) but the resultant force should not be applied less than H/3 above the wall base, where H equals the wall height. We recommend the procedure in AASHTO Section 11 Appendix A11 be used to compute the resultant force location when combining active and dynamic earth pressure (i.e. applied at 0.4H to 0.5H). We also note that equation A11.3.1-1contains an error in AASHTO 2012 and the correct equation is provided in AASHTO 2014.

For the strength limit state, AASHTO Section 11.6.3.6 states that sliding resistance of abutment walls shall be determined in accordance with the provisions for sliding of spread footings in AASHTO Section 10.6.3.4. AASHTO Section 10.6.3.4 refers to AASHTO Table 10.5.5.2.2-1 for strength limit state resistance factors.

0.8 should be applied to the nominal coefficient of sliding resistance value (AASHTO Table 10.5.5.2.2- 10.5.5.3.3) should be applied to the nominal coefficient of sliding resistance value. Load factors have not been included in the Table 7 values.

The static settlement of the proposed 10- to 15-foot high embankments could be 1 to 2 inches and this settlement should occur as the embankment load is applied. The estimated static settlement is based on empirical correlations between soil modulus and SPT blow count.

6.4 Other Retaining Wall Design

A retaining wall will be constructed to create the bike path and trail near the Caltrans Maintenance Station. Boring B-2-15 was drilled near the proposed wall to determine the subsurface conditions behind and below the wall. Recommendations for this wall are provided in the final Truckee River Bridge report.

7.0 CONSTRUCTION AND MATERIALS SPECIFICATION

The applicability of the design parameters in Section 6.0 is contingent on good construction practice. Poor construction techniques may alter conditions from those upon which our recommendations are based, and therefore result in poor performance. Our analyses assumed that this project is constructed according to FP-14. The following sections provide additional construction considerations for this project.

7.1 Driven Pile Installation

Piles shall be driven at the site in general accordance with 2014 Federal Highways Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects (FP-14). The following information is presented for pile installation.

Installation. Based on the subsurface conditions, the piles may be driven with flat bottom plates. When the piles have been driven to the design tip elevation, they would be backfilled with concrete.

Pile-Driving Equipment. Fixed-lead pile driving equipment is generally recommended to drive the proposed piles, although the use of hanging or swinging leads is often used provided that the piles are aligned properly. A combination of vibratory and diesel-powered impact hammers would likely be used for driving the proposed piles. The contractor would provide the vibratory and impact hammer based on their proposed means and methods. We anticipate that an APE Model D 46 or similar may be used to impact the piles. All pile-driving equipment and contractor submittals shall be in accordance with FP-14 specifications.

Pile Driving Conditions. Piles supporting the proposed bridge would be installed through existing fill material, at the abutment locations, and into the very dense clayey gravel soil. Obstructions, such as cobbles, boulders, logs and other large debris, may be encountered during the installation of the piles. Remedial measures should be developed by the contractor to mitigate the impact of obstructions on the pile installation in accordance with FP-14 specifications.

Wave Equation Analysis. Efficient pile driving can be defined as driving the pile to the desired ultimate capacity at a reasonable blow count, say less than 140 blows per foot and as close to the yield strength of the pile material as possible. To establish driving criteria for pile installation, we recommend the Wave Equation Analyses for Pile driving (WEAP) be performed utilizing data for the actual hammer/pile combination to be used in installing the production piles. This method allows evaluation of driving stresses so that an appropriate pile-driving hammer size can be selected to obtain the desired pile capacity with reasonable blow counts and without damaging the piles. This analysis also provides an estimate of the ultimate pile capacity for a given driving resistance. A WEAP analysis and pile driving criteria should be performed after the pile hammer is selected by the contractor.

Test Pile Program. The recommendations for pile foundations and, in particular, the recommendations for pile penetrations and resistance are based on empirical data and subsurface conditions encountered at the site. To substantiate our recommendations and support the resistance factors indicated on Figure 8, we recommend that a test pile program be implemented.

The test pile program would consist of driving indicator piles and performing dynamic pile tests using a Pile-Driving Analyzaer (PDA). We recommend that a minimum of two piles and at least two percent of the production piles be driven as indicator piles. One indicator pile should be driven at each abutment. During the indicator pile driving, we recommend that dynamic measurements, using a PDA, be taken and Case Pile Wave Analysis Program (CAPWAP) be performed on each test pile. Based on our experience, dynamic pile tests are one of the most cost-effective methods for determining the total ultimate resistance and load distribution of the piles. Test piles may be used as production piles if they meet the specified installation procedures and requirements.

7.2 Site Preparation

Subgrade for areas that will receive fill or new pavement should be cleared of trees, shrubs, and any other vegetation; grubbed of stumps and large roots; and stripped of any topsoil and underlying soils that contain significant amounts of roots or other organic matter in accordance with Section 201 and 203 of FP-14. Any rubble or debris should be removed from the subgrade.

This subgrade preparation should extend at least 3 feet beyond the toe of the proposed roadway embankments. After stripping off topsoil, we recommend preparing the subgrade as follows:

Proof-roll the footing and embankment subgrade soil with a heavily loaded dump truck or vibratory roller (or equivalent).

Remove any soft/loose zones noted during rolling and replace those areas with structural fill material constructed in accordance with the recommendations presented below.

Compact the upper 2 feet of the exposed subgrade to at least 95 percent of its maximum dry density determined using AASHTO T-99 Method C.

7.3 Backfill Material, Placement, and Compaction

Prior to the placement of any fill or roadway construction, any standing water should be drained from the area and the exposed subgrade proof-rolled as described in Section 7.2. Embankment fill should be placed on the proof-rolled subgrade material. We recommend using the embankment fill as specified in Section 704.01 of FP-14.

Embankment fill should not contain organics or deleterious material. It should be placed in horizontal lifts and compacted to at least 95 percent of its maximum dry density determined using AASHTO T-99 Method C and the thickness of compacted lifts should not exceed 6 inches.

Trench backfill should be placed in lifts not exceeding 4 inches if compacted with hand-operated equipment, or 8 inches if compacted with heavy equipment. Each lift should be compacted to at least 92 percent of the maximum dry density (AASHTO T-99 Method C) 12 inches or more below the pavement subgrade, and to 95 percent within 12 inches of the pavement subgrade. We recommend a minimum cover over utilities of 2 feet from the crown of the pipes or conduits to the top of the pavement subgrade. Catch basins, utility vaults, and other structures installed flush with the pavement should be designed and constructed to transfer wheel loads to the base of the structure.

7.4 Temporary Groundwater Control

Throughout the proposed excavation areas, the on-site soil contains enough fine-grained material to make it moisture sensitive, and groundwater could be encountered in the excavations;

therefore, control of surface and groundwater would be necessary to maintain the integrity of the material and a firm working platform.

The Contractor should be responsible for the control of groundwater and surface water within the contract limits. In this regard, sloping, slope protection, ditching, sumps, dewatering, and other measures should direct water away from work areas.

7.5 General Excavation, Temporary Cut and Permanent Slopes

To provide safe working conditions and prevent ground loss, excavation slopes should be the responsibility of the Contractor, because the Contractor would be at the job site to observe and control the work. All current and applicable safety regulations regarding excavation slopes and shoring should be followed.

Excavations could be accomplished with conventional excavating equipment, such as a dozer, front-end loader, excavator, or backhoe. For planning purposes, we recommend that temporary, unsupported, open-cut slopes be no steeper than 2 Horizontal to 1 Vertical (2H:1V) in the near-surface soil. Flatter cut slopes could be required where looser soil or seepage zones are encountered. We recommend that all exposed cut slopes be protected with a waterproof covering during periods of wet weather to reduce sloughing and erosion. We recommend that permanent slopes be no steeper than 2H:1V.

Excavated material (or stockpiles of construction materials or equipment) should not be placed closer to the edge of any excavation than the depth of the excavation, unless the excavation is shored and such materials are accounted for as a surcharge load on the shoring system.

7.6 Utilities

Utility trenching should be accomplished in accordance with FP-14. We anticipate that excavation could be accomplished with conventional excavation equipment. If debris is encountered during trench excavation, it may be necessary to increase the thickness of bedding material below utilities to maintain a sufficient thickness above the debris in some areas. Soil exposed at the bottom of the deep trenches may be easily softened or disturbed by construction equipment and operations, especially near the groundwater table. The contractor should take all necessary steps to protect the subgrade from becoming disturbed. If the subgrade is disturbed due to soft or wet conditions, additional soil excavation below the bedding level is recommended. If the subgrade is relatively dry, the excavated soil can be replaced with additional foundation stabilization material. If wet conditions are present in soft compressible soil after over-excavation, then we recommend placing 6- to 12-inch quarry spalls to stabilize the subgrade, before placing foundation stabilization material.

Backfilling of utility trenches should be completed using structural fill compacted as specified above. There should be sufficient cover over the pipe, however, so that when heavy compactors are used, the pipe is not damaged during backfill compaction.

8.0 CONSTRUCTION OBSERVATION

Geotechnical design recommendations are developed from a limited number of explorations and tests. Therefore, recommendations may need to be adjusted in the field. To this end, we recommend that a construction observation and monitoring program be implemented for the project and that Shannon & Wilson be retained to observe the geotechnical aspects of construction. This observation would allow us to confirm that conditions encountered are consistent with those used in our analyses. At the same time, it will allow us to provide expedient recommendations should conditions be revealed during construction that are different from those anticipated.

9.0 LIMITATIONS

This report has been prepared for the exclusive use of CH2M and CFLHD for the purpose of providing geotechnical recommendations for the referenced project. It should be made available to prospective contractors for the purpose of providing information on factual data only, and not as a warranty of subsurface conditions, such as those interpreted from the logs and discussions of subsurface conditions in this report.

The analyses, conclusions, and recommendations contained in this report are based on site conditions as they existed at the time of our explorations. If there is a substantial lapse of time between the submission of this report and the start of construction at the site, if alternative design codes are adopted, or if conditions change due to natural causes or construction operations at or adjacent to the site or appear different from those described in our report, we recommend that we be retained to review our report to determine the applicability of the conclusions and recommendations considering the changed conditions and time lapse.

Caltrans provided review comments on the geotechnical reports submitted as the project progressed. Comments on the report as submitted by Caltrans including our responses are included in Appendix C.

Within the limitations of scope, schedule and budget, the analyses, conclusions and recommendations presented in this report were prepared in accordance with generally accepted professional geotechnical and geological principles and practice in this area at the time this report was prepared. We make no other warranty, either express or implied.

Unanticipated soil conditions are…