B09_A1_Attach_17_Geotechnical_Report_Ahwahnee_Hotel_FINAL_0001.pdf

GEOTECHNICAL INVESTIGATION

AHWAHNEE HOTEL

Yosemite National Park, California

Delaware North Companies San Francisco, California

24 November 2009 Project No. 4994.01

ENVIRONMENTAL AND GEOTECHNICAL CONSULTANTS

555 MONTGOMERY STREET, SUITE 1300 SAN FRANCISCO CALIFORNIA 94111 T 415 955 9040 F 415 955 9041 www.treadwellrollo.com

Project 4994.01

Mr. Brian Fulce

Facility Services P.O. Box 578 9032 Village Drive Yosemite, California 95389

Subject: Geotechnical Investigation Ahwahnee Hotel Rehabilitation Project Yosemite National Park, California

Dear Mr. Fulce:

This letter transmits our geotechnical investigation report for the proposed rehabilitation of the Ahwahnee Hotel in Yosemite National Park, California. Additional copies have been distributed to the team members listed on the last page of the report. This cover letter omits detailed findings and conclusions; therefore, anyone relying on the report should read it in its entirety. Our conclusions and recommendations apply only to the project described in the report.

The Ahwahnee Hotel is the centerpiece of the Ahwahnee National Historic Landmark, which is located at the east end of Yosemite Valley in Yosemite National Park, and includes adjacent guest cottages, employee dormitories, a stone entrance gate, and other improvements such as a swimming pool and an outdoor plaza. The hotel is a multi-level structure ranging from three to six stories, with a footprint of approximately 40,000 square feet. The hotel was constructed in the mid-1920’s and has been in continuous service since that time.

We understand the rehabilitation project will include renovations and upgrades to the Ahwahnee Hotel, adjacent cottages, employee dormitory, support buildings, landscaping, entrance gate, and stone porte-cochere. Our investigation and studies address the improvements to the Ahwahnee Hotel only. Seismic upgrades to the hotel will likely include new shear walls, which will require new foundations.

The results of our investigation indicate the Ahwahnee Hotel site is underlain primarily by alluvial deposits, consisting of sand and gravel. During our investigation we encountered sand, silty sand, and sand with gravel deposits ranging in consistency from loose to very dense. In general, the sand layers encountered are loose to medium dense in the upper 10 to 15 feet and become denser with depth;

however, loose sand deposits were encountered at various depths between 33 and 58 feet. Groundwater was encountered during our investigation at depths of between approximately 10-1/2 and 13-1/2 feet below ground surface. Groundwater is seasonally variable and fluctuations of several feet from those measured during our investigation should be anticipated.

Depending upon the design level of earthquake shaking considered, the combination of loose to medium dense sand and high groundwater level could result in liquefaction-induced settlement.

San Francisco, California

Project No. 4994.01

TABLE OF CONTENTS

1.0 INTRODUCTION

2.0 PROJECT DESCRIPTION

3.0 SCOPE OF SERVICES

4.0 FIELD EXPLORATION AND LABORATORY TESTING

4.1 Borings

4.2 In-Situ Shear Wave Velocity Measurement

5.0 LABORATORY

6.0 GEOLOGY AND SITE SUBSURFACE CONDITIONS

6.1 Regional Geology

6.2 Yosemite Valley Geology

6.3 Site Subsurface Conditions

7.0 SEISMIC CONSIDERATIONS

7.1 Regional Seismicity and Faulting

7.2 Geologic Hazards

7.2.1 Liquefaction and Lateral Spreading

7.2.2 Ground Rupture

8.0 DISCUSSION AND CONCLUSIONS

8.1 Foundations

9.0 RECOMMENDATIONS

9.1 Site Preparation, Fill Compaction, Grading

9.2 Foundation

9.3 Site-Specific Ground Motion Analysis

9.4 ASCE Standard ASCE/SEI 41-06 Code Seismic Design Values

10.0 GEOTECHNICAL SERVICES DURING CONSTRUCTION

11.0 LIMITATIONS

REFERENCES

FIGURES

APPENDIX A – Boring Logs

APPENDIX B – Shear Wave Velocity Profile

APPENDIX C – Laboratory Test Data

APPENDIX D – Site-Specific Ground Motions

DISTRIBUTION

LIST OF FIGURES

Figure 1 Site Location Map

Figure 2 Site Plan

Figure 3 Map of Major Faults and Earthquake Epicenters in the Site Vicinity

Figure 4 Recommended Spectra

APPENDIX A

Boring Logs

Figures A-1 Logs of Borings TR-1 through TR-3 through A-3

Figure A-4 Soil Classification Chart

APPENDIX B

Shear Wave Velocity Profile

Figures B-1 Shear Wave Velocity Profile – Boring TR-1

APPENDIX C

Laboratory Test Data

Figures C-1 Particle Size Analysis through C-4

Yosemite National Park, California

1.0 INTRODUCTION

This report presents the results of the geotechnical investigation performed by Treadwell & Rollo, Inc. for the proposed rehabilitation of the Ahwahnee Hotel in Yosemite National Park, California. The site is the valley of Yosemite National Park, approximately ¼ mile north of Highway 140, as shown on Figure 1.

2.0 PROJECT DESCRIPTION

The Ahwahnee Hotel is the centerpiece of the Ahwahnee National Historic Landmark, which includes adjacent guest cottages, employee dormitories, a stone entrance gate, and other improvements such as a swimming pool and an outdoor plaza. The hotel is a multi-level structure ranging from three to six stories, with a footprint of approximately 40,000 square feet. The hotel was constructed in the mid-

1920’s and has been in continuous service since that time.

Previous studies of the Ahwahnee Landmark have identified outdated, inefficient, and non-code compliant structural systems in the various structures. We understand the rehabilitation project will include seismic strengthening, rehabilitation and upgrades to the Ahwahnee Hotel, adjacent cottages, employee dormitory, support buildings, landscaping, entrance gate, and stone porte-cochere. Seismic improvements will likely include construction of new shear walls. This report addresses the improvements to the Ahwahnee Hotel only.

3.0 SCOPE OF SERVICES

Our scope of services for the current investigation was described in our proposal and proposal amendment, dated 14 July 2009 and 17 July 2009, respectively.

The primary objective of our investigation was to evaluate the potential seismic hazards at the project site, including developing site-specific response spectra and time histories, and evaluating the potential

49940102.CER 24 November 2009 for liquefaction and liquefaction-induced settlements. On the basis of the results of our investigation, laboratory results and engineering analysis, we developed conclusions and recommendations regarding:

• soil and groundwater conditions

• site seismicity and seismic hazards, including potential for fault rupture, ground shaking, liquefaction, lateral spreading and seismically-induced settlements

• site-specific response spectra per definitions in ASCE 41-06

• seven pairs of earthquake time histories

• appropriate foundation types for the existing structure and new shear walls

• estimates of foundation settlement

• design parameters for appropriate foundation types(s)

• passive earth pressures for resistance of lateral loads

• site grading and excavation, including criteria for fill quality and compaction

• site class and seismic coefficients per the appropriate building code

• construction considerations

4.0 FIELD EXPLORATION AND LABORATORY TESTING

Prior to performing our field investigation, we reviewed the results of previous studies (Kleinfelder 2000 and 2009) at and in the vicinity of the site. Our field exploration, which was performed on 4 through 6

August 2009, consisted of drilling three borings, identified as Borings TR-1 through TR-3, and performing one in-situ shear wave velocity measurement in Boring TR-1. The in-situ shear wave velocity measurements assisted us in our evaluation of liquefaction potential, as well as classifying the site seismic soil classification. Details of each aspect of our field investigation are provided in the following sections.

4.1 Borings

Prior to commencing our field work, we coordinated our activities with Delaware North Companies and

Yosemite National Park staff to minimize disturbance to the park and hotel operations. We also notified

Underground Service Alert (USA) to locate utilities in the vicinity of the borings.

Boring TR-1 was drilled to a depth of 101.5 feet below existing grade and Borings TR-2 and TR-3 were drilled to a depth of 51.5 feet. The borings were advanced using truck-mounted, rotary wash drilling equipment operated by Pitcher Drilling of Palo Alto, California. During drilling, our field engineer logged the borings and obtained samples of the material encountered for visual classification and laboratory testing. Logs of the borings are presented in Appendix A as Figures A-1 through A-3. The material encountered was classified according to the soil classification system described on Figure A-4. The logs of borings by others and the associated soil classification system that we reviewed are also presented in

Appendix A.

Soil samples were obtained using the following samplers:

• Sprague & Henwood (S&H) split-barrel sampler with a 3.0-inch outside diameter and 2.5-inch inside diameter, lined with 2.43-inch inside diameter brass tubes

• Standard Penetration Test (SPT) sampler with a 2.0-inch outside diameter and 1.5-inch inside diameter, lined with 1.375-inch inside diameter stainless steel tubes

S&H and SPT samplers were driven with a 140-pound, above-ground automatic safety hammer falling

30 inches. The blow counts required to drive the S&H and SPT samplers S&H sampler the final 12 inches of an 18-inch drive were corrected using a factor of 1.2 and 0.8, respectively, to approximate SPT (N60) blow counts and are shown on the boring logs.

The liners for the soil samples were sealed at each end to retain field moisture content. All samples were appropriately labeled, including designation, sample number and depth, date, and Treadwell & Rollo project number.

Upon completion of drilling, the borings were backfilled with cement grout. All drilling spoils were placed in 55-gallon drums and stored at an off-site location. After chemical profiling of the contents, we coordinated disposal of the drums.

4.2 In-Situ Shear Wave Velocity Measurement

In-situ shear wave velocity measurement was performed by GEOVision Geophysical Services of Corona, California on 5 August 2009 using suspension P-S velocity logging in Boring TR-1. The P-S logging system uses a 7-meter probe, containing a source and two receivers spaced one meter apart, suspended by a cable. An armored 4-conductor cable is used both to support the probe and to convey data to and from a recording/control device on the surface. The probe was lowered into the borehole to a specified depth (a rotary encoder on the winch measured probe depth), where the source generated a pressure wave in the borehole fluid. The pressure wave was converted to seismic waves (P and S) at the borehole wall. Along the wall at each receiver location, the P and S waves were converted back to pressure waves in the fluid and received by the geophones, which sent the data to the recorder on the surface. The elapsed time between arrivals of the waves at the receivers was used to determine the average velocity of a one-meter-high column of soil around the borehole. The shear wave velocity profile measured in

Boring TR-1 is presented on Figure B-1 in Appendix B.

5.0 LABORATORY

Samples were reviewed at our office and representative soil samples were selected for laboratory testing.

Samples were tested to measure moisture content, dry density, particle size, and fines content.

Laboratory test results are presented in Appendix C and summarized on the boring logs.

6.0 GEOLOGY AND SITE SUBSURFACE CONDITIONS

6.1 Regional Geology

Yosemite National Park occupies approximately 1,170 square miles within the central portion of the Sierra

Nevada. Granitic bedrock is widespread in Yosemite National Park and dominates a significant portion of the Sierra Nevada. The granitic rock formed deep within the earth as a mass of melted rock that slowly cooled beneath the ground surface, eventually creating subsurface bodies of solidified granitic rock called the Sierra Nevada batholith.

Between 100 million and 65 million years ago, magma formation slowed and a long period of erosion began in the Sierra Nevada. Erosion removed the overlying rocks and exposed the underlying core of the batholith. About 15 million years ago, the Sierra Nevada in the Yosemite region had gently rolling upland topography and a much lower elevation than the present-day range. The Merced River flowed westward at a gentle gradient through a broad river valley. Volcanic activity, prevalent in the northern Sierra

Nevada, deposited ash, filled valleys, buried streams, and altered river courses.

Later, mountain-building activity was reactivated and tilted the Sierra Nevada to form its relatively gentle western slope and the more dramatic, steep eastern slopes. The uplift increased the gradients of the rivers and resulted in deeply incised river valleys. Between 2 and 3 million years ago, snow and ice accumulated as glaciers at the higher alpine elevations began to move westward down the mountain valleys. The downslope movement of the glaciers cut and sculpted the valleys, cirques, and other glacially formed landforms throughout the Yosemite region and the Sierra Nevada.

6.2 Yosemite Valley Geology

Yosemite Valley trends approximately east-west, is about 6.8 miles long, and varies from about ½- to

¾-mile wide. The Yosemite Valley forms a classic, glaciated, U-shaped valley, providing text-book examples of many glacial erosion features. The Merced River flows through the valley at elevations ranging from 4,200 to 3,800 feet above mean sea level on sediments deposited by the glaciers as well as lacustrine (lake bottom) and landslide deposits accumulated since glacial recession approximately

15,000 years ago. The glaciers carved Yosemite Valley out of primarily granitic rock leaving valley walls that tower 1,500 to 4,000 feet above the essentially flat valley floor.

The downslope movement of the glaciers cut and sculpted the U-shaped valley that is present today.

When glaciers melt, the rock debris they transport (glacial till) is deposited in landforms known as moraines. Two prominent moraines were formed in Yosemite Valley after the last glacier retreated about

15,000 years ago. The terminal moraine, marking the furthest extent of the glacier, lies just east of

Bridalveil Meadow. The El Capitan moraine, lying further east, is a recessional moraine, formed as the glacier receded up the Valley from its furthest extent. As the last ice melted, thick moraine deposits dammed the heavy melt water flow and created prehistoric Lake Yosemite. Ultimate breaching of the moraine dam and continued stream deposition gradually filled Lake Yosemite. Seismic refraction studies

(Gutenberg et al. 1956) documented about 2,000 feet of valley fill at the head of the Valley between the

Ahwahnee Hotel and Curry Village.

6.3 Site Subsurface Conditions

The results of our investigation indicate the Ahwahnee Hotel site is underlain primarily by alluvial deposits, consisting of sand and gravel. During our investigation we encountered sand, silty sand, and sand with gravel deposits ranging in consistency from loose to very dense. In general, the sand layers encountered are loose to medium dense in the upper 10 to 15 feet and become denser with depth;

however, loose sand deposits were encountered in Boring TR-1 at depths of approximately 33 to 40 feet and 44 to 58 feet and in Boring TR-3 at a depth of approximately 33 to 48 feet. The gravel layers encountered range in thickness of between 4-1/2 feet to 11 feet and are dense to very dense.

During our investigation, groundwater was encountered at depths of between approximately 10-1/2 and

13-1/2 feet below ground surface. Groundwater is seasonally variable. During the winter months, groundwater, within the valley, could be several feet higher than depths measured during our investigation. Likewise, during periods of low flow within the Merced River, groundwater could be several feet lower than those depths measured during our investigation.

7.0 SEISMIC CONSIDERATIONS

7.1 Regional Seismicity and Faulting

Earthquakes felt in the Yosemite Valley usually have epicenters in western California or along the eastern flank of the Sierra Nevada. Historically, nine earthquakes (not including aftershocks) have been felt and reported in the Yosemite Valley. Four of these events have triggered rockfall events (Wieczorek and

Jaeger 1996). No active or potentially active earthquake faults have been identified in the mountain region of Yosemite National Park (Hart 1990).

Earthquakes affecting Yosemite Valley have been generated on faults that lie west and east of the Sierra

Nevada. West of the Sierra Nevada seismogenic sources include the Foothills fault zone, various faults along the western boundary of the Great Valley (Great Valley Thrust Zone) and the eastern boundary of the San Andreas Fault zone. To the east of the Sierra Nevada lies the Eastern Escarpment with associated seismogenic geological structures, including the Mono Craters – Long Valley Caldera complex, and the escarpment bounding faults primarily within the Owens Valley fault zone, and in the Basin and

Range Province to the east of the Sierra Nevada (CDMG 1996).

The Foothills fault zone extends for approximately 220 miles in a north-south direction within the foothills of the Sierra Nevada, approximately 50 miles west of Yosemite Valley. This fault zone is capable of generating 6.5 maximum moment magnitude1 earthquake events (Petersen et al. 1996). One significant historical event of magnitude 5.7 occurred in 1975 near Oroville (CDMG 1975).

The Basin and Range province lies east of the Sierra Nevada and is the source of significant earthquake activity. The Mono Lake fault is characterized as active (Petersen et al. 1996) and capable of a 6.6 maximum moment magnitude event and is located approximately 35 miles northeast of Yosemite Valley within the Mono Craters – Long Valley Caldera region. In October 1990, the Mono Lake fault experienced a 5.7 magnitude event. This earthquake was felt as far west as Sacramento and the San Francisco Bay

Area and caused landslides and rockfalls at Tioga Pass and on the Big Oak Flat Road (McNutt et al.

1991). In May of 1980 there was a sequence of four earthquakes measuring 6.0 to 6.2 at Mammoth

Lakes that triggered nine rock slides and falls at Yosemite Valley approximately 44 miles west of the epicenters (Wieczorek and Jaeger 1996).

The 1872 Owens Valley earthquake, the epicenter of which was located approximately 112 miles southeast of Yosemite Valley, triggered five reported rock falls (Wieczorek and Jaeger, 1996). The

Owens Valley earthquake is estimated to have had a magnitude between 7.6 and 8.0 and was one of the largest earthquakes in U.S. history (Goter 1992 and Goter et al. 1994).

The nearest known active fault is the Hartley Springs Fault, which is 45 kilometers from the site. This and other faults of the region are shown on Figure 3. For each of the active faults, the distance from the site and estimated mean characteristic Moment magnitude2 [2007 Working Group on California

Earthquake Probabilities (WGCEP) (2007) and Cao et al. (2003)] are summarized in Table 1.

1 Moment magnitude is an energy-based scale and provides a physically meaningful measure of the size of a faulting event. Moment magnitude is directly related to average slip and fault rupture area.

2 Moment magnitude is an energy-based scale and provides a physically meaningful measure of the size of a faulting event. Moment magnitude is directly related to average slip and fault rupture area.

TABLE 1

Regional Faults and Seismicity

Fault

Approximate

Distance From Site (kilometers)

Direction From Site

Mean Characteristic

Moment Magnitude

Hartley Springs 45 East 6.6 Mono Lake 46 East 6.6 Robinson Creek 56 Northeast 6.4 Hilton Creek 61 East 6.7 Round Valley 76 East 6.8

Figure 3 also shows the earthquake epicenters for events with magnitude greater than 5.0 from

January 1800 through December 2000.

7.2 Geologic Hazards

During a major earthquake on a segment of one of the nearby faults, moderate shaking is expected to occur at the site. Moderate to strong shaking during an earthquake can result in ground failure such as that associated with soil liquefaction3, lateral spreading4 and differential compaction5. Each of these hazards is discussed in the following subsections.

7.2.1 Liquefaction and Lateral Spreading

Flow failure, lateral spreading, differential settlement, loss of bearing strength, ground fissures, and sand boils are evidence of excess pore pressure generation and liquefaction.

3 Liquefaction is a phenomenon in which saturated, cohesionless soil experiences a temporary loss of strength due to the buildup of excess pore water pressure, especially during cyclic loading such as that induced by earthquakes. Soil most susceptible to liquefaction is loose, clean, saturated, uniformly graded, fine-grained sand and silt of low plasticity that is relatively free of clay.

4 Lateral spreading is a phenomenon in which surficial soil displaces along a shear zone that has formed within an underlying liquefied layer. Upon reaching mobilization, the surficial blocks are transported downslope or in the direction of a free face by earthquake and gravitational forces.

5 Differential compaction is a phenomenon in which non-saturated, cohesionless soil is compacted by earthquake vibrations, causing differential settlement.

We performed our liquefaction analysis in accordance with the State of California Special Publication 117, Guidelines for Evaluation and Mitigation of Seismic Hazards in California and following the procedures presented by Idriss and Boulanger (2008) and Andrus and Stokoe (2000).

In simple terms, these analyses methods calculate a factor of safety against liquefaction triggering by taking the ratio of soil strength (resistance of the soil to cyclic shaking) to the seismic demand

(seismically-induced stresses) that can be expected from a design level seismic event. Specifically, two distinct terms are used in the liquefaction triggering analyses:

• Cyclic Resistance Ratio (CRR), which quantifies the soil’s resistance to cyclic shaking; a function of soil depth, density, depth of groundwater, earthquake magnitude, and overall soil behavior

• Cyclic Stress Ratio (CSR), which quantifies the stresses that may develop during cyclic shaking

The factor of safety (FS) against liquefaction triggering can be expressed as the ratio of CRR over CSR.

For our analyses, if the FS for a soil layer is less than 1.3, it is considered possible that the soil layer may liquefy and generate excess pore water pressure during a large seismic event. We calculated liquefaction-induced settlements in cases where the computed FS within the depth of interest was less than 1.3.

CRR is a function of the level of shaking and depends on SPT penetration resistance blow counts normalized for overburden pressure and corrected for fines content. The CSR is based on the relative density of the soil, total and effective overburden pressures, the estimated peak horizontal acceleration at the ground surface (amax), and a stress reduction coefficient (rd).

For our analyses, we considered two levels of ground shaking: Basic Safety Earthquake 1 and 2 (BSE-1 and BSE-2, see Section 9.3 for further details). BSE-1 and BSE-2 are probabilistic seismic hazard spectra having 10 and 2 percent probabilities of exceedance in 50 years, respectively. The annual frequencies associated with BSE-1 and BSE-2 are about 0.0021 and 0.0004, respectively. The site-specific peak ground accelerations (PGA’s) associated with the BSE-1 and BSE-2 levels are 0.131g and 0.218g, respectively. Using these site-specific PGA’s, we conclude there is no potential for liquefaction at the site under the BSE-1 level of shaking; however, for the BSE-2 level of shaking, it appears several loose sand layers encountered in Boring TR-1 at depths of approximately 33 to 40 feet and 44 to 58 feet and in

Boring TR-3 at a depth of approximately 33 to 48 feet could potentially liquefy during a BSE-2 level of shaking event resulting in liquefaction-induced settlement. No potentially liquefiable soil was identified in

Boring TR-2 under either BSE-1 or BSE-2 level of shaking. It appears the looser sandy deposits tend to exist closer to Merced River. Furthermore, it appears the thickness of these layers also increases in a southerly direction towards Merced River.

We estimate the PGA to induce liquefaction at the site is approximately 0.158g. This corresponds to an

800-year return period and an annual frequency of approximately 0.00125. It should be noted that annual frequencies associated with the 800-year return period and BSE-2 levels of shaking are considered rare to very rare seismic events.

Using the Tokimatsu and Seed (1987) method for evaluating earthquake-induced liquefaction settlement, we calculate settlement of 0 to 11 inches could occur in the free-field, i.e. beyond the building footprint, during the BSE-2 level of shaking. The available methods, including Tokimatsu and Seed (1987), currently used for evaluation of liquefaction-induced settlements are intended to be used for free-field evaluations and do not consider the effects of a building or other structure. Studies by Rollins and Seed

(1990), Liu and Dobry (1997), and Dashti et al. (2009) have shown that liquefaction-induced settlements beneath buildings may not be as large as those estimated for the free-field because of the building’s influence on the suppression of generation of excess pore water pressures beneath structures.

Furthermore, these studies have concluded that liquefaction-induced settlements beneath buildings are a function of building width. Using the results of these studies, we estimate 0 to 6 inches of liquefaction-induced settlement beneath the building footprint during the BSE-2 level of shaking. The variation in magnitude of settlement is due to the variation in thicknesses and relative density of the sandy material encountered in our borings. For the estimated settlement beneath the building, we estimate differential settlement on the order of one inch over a distance of about 40 feet. Table 2 presents estimates of liquefaction-induced settlements in the free-field and beneath the building in the vicinity of Borings TR-1 through TR-3. To estimate differential settlement in a specific area of the building, we recommend linearly interpolating between the settlement estimates presented in Table 2.

A summary of our boring data and associated potential liquefaction-induced settlement under BSE-2 level of shaking in the free-field and beneath the structure, are presented in Table 2.

TABLE 2

Estimate of Liquefaction-Induced Settlement under BSE-2 Level of Shaking

Estimated Liquefaction-Induced Settlement under BSE-2 Level

Shaking (inches)

Boring Number

Free-Field (outside of building)

Beneath Building

TR-1 11 6

TR-2 0 0

TR-3 7 4

The nearest free face slope is the bank of the Merced River, which is approximately 800 feet from the site. The bed of the Merced River near the Ahwahnee Hotel is less than 10 to 15 feet below the ground surface grades at the site. Because the potentially liquefiable (under BSE-2 level of shaking) layers beneath the site appear to be relatively deep (below a depth of about 33 feet) and below the bed of the

Merced River, and the site is relatively flat, we conclude that the potential for lateral spreading at the site is low.

Cyclic densification of non-saturated loose to medium dense sand by earthquake shaking can cause ground surface settlement (differential compaction). Loose to medium dense sand and silty sand layers were encountered above the groundwater in all three borings drilled for this investigation. For both the

BSE-1 and BSE-2 levels of ground shaking we conclude these layers are sufficiently dense so that the potential for cyclic densification and associated settlement is low.

7.2.2 Ground Rupture

Historically, ground surface ruptures closely follow the trace of geologically young faults. The site is not within an Earthquake Fault Zone, as defined by the Alquist-Priolo Earthquake Fault Zoning Act and no known active or potentially active faults exist on the site. Therefore, we conclude the risk of fault offset at the site from a known active fault is low. In a seismically active area, the remote possibility exists for future faulting in areas where no faults previously existed; however, we conclude the risk of surface faulting and consequent secondary ground failure is low.

8.0 DISCUSSION AND CONCLUSIONS

We conclude, from a geotechnical standpoint, that the proposed seismic strengthening and rehabilitation can be constructed as planned. The primary geotechnical issues to be addressed for the project are potential settlement due to liquefaction-induced settlement during a large earthquake (BSE-2 level of ground shaking) and selecting the most appropriate foundation system for the proposed structure. Our conclusions regarding these and other issues are discussed in the remainder of this section.

8.1 Foundations

The architectural plans6 for the existing hotel indicate it is supported on a shallow foundation system consisting of a combination of perimeter continuous footings and interior isolated spread footings. We anticipate medium dense sand will be present at the foundation level. The planned concrete shear walls and other new seismic elements and improvements can be supported on a new shallow foundation system bearing in the medium dense sand.

We judge that settlement of the building is complete under the current building loads. The anticipated structural loads supported on new foundations will cause compression of the sand; the magnitude of total and differential settlements will depend on the new loads and the footing size. Table 3 in Section 9.2 presents a range of bearing pressures and associated anticipated total settlements. Differential settlements should be approximately half of those presented in Table 3. The majority of this settlement should occur during construction. Additionally, we estimate that between 0 and 11 inches of total liquefaction-induced settlement could occur should an earthquake occur resulting in BSE-2 level of shaking. We estimate that seismically-induced differential settlement over a distance of about 40 feet could be on the order of one inch.

9.0 RECOMMENDATIONS

We recommend the proposed concrete shear walls and other new seismic elements and improvements be supported on a shallow foundation system bearing in the medium dense sand. Recommendations for the foundation design and other geotechnical aspects of the project design and construction are presented in the remainder of this section.

6 Architectural plans by Gilbert Stanley Underwood and Co., “A Hotel for the Yosemite Park and Curry Company, Yosemite National Park, California”, dated September 1926.

9.1 Site Preparation, Fill Compaction, Grading

We understand there is a crawl space beneath the floor slab of the Ahwahnee hotel and no new building slab-on-grade is planned as part of the current rehabilitation project. However, we further understand some exterior hardscape/slabs-on-grade may be removed and replaced. Grading operations should commence after demolition and removal of any existing pavements, foundation slabs, utilities, and underground obstructions within the development area. Following demolition, all areas to receive improvements should be stripped of vegetation and organic topsoil, including roots larger than one-inch in diameter. The pavement material, including asphalt, may be segregated from organic topsoil and used as compacted fill, provided it meets the fill requirements presented in this section and is acceptable from an environmental standpoint. The stripped organic soil can be stockpiled for later use in landscaped areas, if approved by the architect; organic topsoil should not be used as compacted fill.

The surface exposed by stripping should be scarified to a depth of at least six inches, moisture-conditioned to near the optimum moisture content, and compacted to at least 90 percent relative compaction7. The exposed ground surface should be kept moist during subgrade preparation.

New fill and backfill should consist of either on-site or imported soil that is non-hazardous, non-corrosive, free of organic matter, has no particles larger than three inches in greatest dimension, has a liquid limit less than 40 and a plasticity index less than 12, and is approved by the geotechnical engineer. In general, the existing near-surface soil is expected to meet the criteria for select fill. A sample of proposed new fill(s) should be submitted to the geotechnical engineer for testing at least three business days prior to use at the site.

Fill and backfill should be placed in horizontal layers not exceeding eight inches in loose thickness, moisture-conditioned to near the optimum moisture content, and compacted to at least 90 percent relative compaction. The upper six inches of the soil subgrade in pavement areas should be compacted to at least 95 percent relative compaction. Fill deeper than five feet should be compacted to at least 95 percent relative compaction. On-site or imported sand with less than 10 percent fines used as fill or backfill should be compacted to at least 95 percent relative compaction.

7 Relative compaction refers to the in-place dry density of soil expressed as a percentage of the maximum dry density of the same material, as determined by the ASTM D1557-07 laboratory compaction procedure.

The surface at the base of the footing excavations and new slabs-on-grade and pavements should be checked by our field engineer. If soft or disturbed material is encountered at the bottom of the excavation, the weak soil should be removed and replaced with properly compacted fill (engineered fill) or lean concrete.

9.2 Foundation

As previously discussed, we anticipate medium dense sand will be exposed at the foundation level.

Shallow foundations consisting of individual spread footings and/or continuous footings may be designed using the bearing pressures shown in Table 3, provided the indicated associated settlements are acceptable.

TABLE 3

Bearing Pressures (Dead Plus Live Loads) and Associated Anticipated Static Settlement

Bearing Pressure (pounds per square foot, psf, dead plus live loads)

Associated Anticipated Static Settlement

(inches)

3,000 1

6,000 2

10,000 3

12,000 4

The bearing pressures shown in Table 3 may be increased by one-third for total loads, including wind or seismic forces. To design the foundation using the modulus of subgrade reaction method, we recommend using a modulus of subgrade reaction of 35 kips per cubic foot (kcf) for static conditions and

45 kcf for dynamic conditions. The moduli are representative of the anticipated settlement under the building loads. After the foundation analysis is completed, we should review the computed settlement and bearing pressure profiles to check that the moduli values are appropriate.

Lateral loads can be resisted by a combination of passive pressure acting on the vertical faces of the mat and friction along the base of the footings. Passive resistance may be calculated using an equivalent fluid weight (triangular distribution) of 250 pounds per cubic foot (pcf). The upper one foot of soil should be ignored unless it is confined by slabs. A base friction value of 0.35 times the dead load of the column load should be used for design. The passive resistance and base friction coefficient values include a factor of safety of at least 1.5.

9.3 Site-Specific Ground Motion Analysis

We performed probabilistic seismic hazard analysis (PSHA) to develop site-specific response spectra for the site. Based on the results of the shear wave velocity testing performed as part our field investigation, the average shear wave velocity in the top 100 feet (30 meters), Vs30, of the site is 226 m/s (740 ft/sec).

As such, the site is classified as SD. We used the computer software EZFRISK 7.34 (Risk Engineering

2009) and Next Generation Attenuation Relationships (NGA) to develop site-specific spectra. The details of our analysis are presented in Appendix D.

We understand the provision of ASCE 41-06 will be used for seismic evaluation and design of the hotel.

As such, we developed response spectra corresponding to BSE-1 and BSE-2. For the site, BSE-2 is defined as a level of ground shaking corresponding to 2 percent probability of exceedance in 50 years

(2,475 year return period) per section C1.6.1.1 and BSE-1 is defined as the lesser of 2/3 times BSE-2 or

10 percent probability of exceedance in 50 years (475 year return period). Furthermore, we used the minimum spectral amplitude criteria presented in Section 1.6.2.1.2 of ASCE 41-06 in the development of recommended site-specific spectra. This provision mandates that the site-specific spectrum shall not be less than 70 percent of General Response Spectrum. Figure 4 presents the recommended spectra for

BSE-1 and BSE-2. The digitized values of these spectra for a damping ratio of five percent are presented in Table 4.

TABLE 4

Recommended Acceleration Response Spectra

Damping Ratio of 5 percent

Period (seconds) BSE-1

Period (seconds) BSE-2

0.01 0.145 0.01 0.235

0.09 0.364 0.09 0.588

0.20 0.364 0.20 0.588

0.30 0.364 0.30 0.588

0.46 0.364 0.45 0.588

0.50 0.336 0.50 0.524

0.60 0.280 0.60 0.442

0.70 0.253 0.70 0.410

0.75 0.245 0.75 0.396

0.80 0.234 0.80 0.380

0.90 0.215 0.90 0.351

1.00 0.200 1.00 0.327

1.50 0.139 1.50 0.233

2.00 0.108 2.00 0.183

3.00 0.068 3.00 0.117

4.00 0.048 4.00 0.086

9.4 ASCE Standard ASCE/SEI 41-06 Code Seismic Design Values

For seismic design in accordance with the provisions of ASCE Standard ASCE/SEI 41-06 code we recommend the following:

• BSE-1 Level of Shaking Ss and S1 of 0.34g and 0.10g, respectively (USGS 2008 Mapped Values)

• BSE-1 Level of Shaking Site Coefficients FA and Fv of 1.528 and 2.40, respectively

• BSE-2 Level of Shaking Ss and S1 of 0.66g and 0.18g, respectively (USGS 2008 Mapped Values)

• BSE-2 Level of Shaking Site Coefficients FA and Fv of 1.272 and 2.08, respectively

10.0 GEOTECHNICAL SERVICES DURING CONSTRUCTION

We should review the final project plans and specifications to check that they are in general conformance with the intent of our recommendations. During construction, our field engineer should provide on-site observation and testing during site grading, placement and compaction of backfill, and construction of foundations. These observations will allow us to compare actual with anticipated soil conditions and check that the contractor’s work conforms to the geotechnical aspects of the plans and specifications.

11.0 LIMITATIONS

The conclusions presented in this report apply to the site and construction conditions as we have described them and are the results of engineering studies and our interpretations of the existing geotechnical conditions. Actual conditions may vary. Should conditions differ substantially from those we anticipate, some modifications to our conclusions and recommendations may be necessary.

REFERENCES

Abrahamson, N. A, and Silva, W. J. (2008). “Summary of Abrahamson & Silva NGA Ground-Motion Relations.” Earthquake Spectra, 24(1), 67-97.

Andrus, R. D., and Stokoe, K. H. (2000). “Liquefaction Resistance of Soils from Shear-Wave Velocity.”

Journal of Geotechnical and Geoenvironmental Engineering, ASCE 126(11), 1015-025.

ASCE Standard ASCE/SEI 41-06.

Boore, D. M. and Atkinson, G. M. (2008). “Ground-Motion Prediction Equations for the Average Horizontal Component of PGA, PGV, and 5%-Damped PSA at Spectral Periods between 0.01 s and 10.0 s.”

Earthquake Spectra, 24(1), 99-138.

California Division of Mines and Geology (1996). “Probabilistic Seismic Hazard Assessment for The State Of California.” DMG Open-File Report 96-08.

Campbell, K. W. and Bozorgnia, Y. (2008). “NGA Ground Motion Model for the Geometric Mean Horizontal Component of PGA, PGV, PGD, and 5%-Damped PSA at Spectral Periods between 0.01 s and

10.0 s.” Earthquake Spectra, 24(1), 139-171.

Cao, T., Bryant, W. A., Rowshandel, B., Branum D. and Wills, C. J. (2003). “The Revised 2002 California Probabilistic Seismic Hazard Maps.”

Chiou, B. S.-J., and Youngs, R. R. (2008). “An NGA Model for the Average Horizontal Component of Peak Ground Motion and Response Spectra.” Earthquake Spectra, 24(1), 173-215.

Cornell, C. A. (1968). “Engineering Seismic Risk Analysis.” Bulletin of the Seismological Society of America, 58(5).

Dashti, S.M., Bray, J.D., Pestana, J.M., Riemer, M., Wilson, D. (2009, Accepted Manuscript, Not Yet Published), “Mechanisms of Seismically-Induced Settlement of Buildings with Shallow Foundations on Liquefiable Soil.” Journal of Geotechnical and Geoenvironmental Engineering, Manuscript submitted 28 June 2008, Accepted 9 June 2009, Posted ahead of print, 20 June 2009.

Goter, S. G., (1992), “Southern California Earthquakes,” USGS Open File Report 92-533.

Goter, S. G., Oppenheimer, D. H., Mori, J. J., Savage, M. K., and Masse, R.P. (1994), “Earthquakes in California and Nevada,” USGS Open File Report 94-647, scale 1:1,000,000, 1 sheet.

Gutenberg, B., Buwalda, J. P., and Sharp, R. P., (1956), “Seismic Explorations on the Floor of Yosemite Valley, California”, GSA Bulletin; August 1956; v. 67; no. 8; p. 1051-1078;

Hart, E. W. (1990) “Fault-Rupture Hazard Zones in California,” California Division of Mines and Geology, Special Publication 42, Revised 1990.

Idriss, I. M. and Boulanger. R. W. (2008). “Soil Liquefaction During Earthquakes.” Earthquake Engineering Research Institute.

Kleinfelder (2000). “Geotechnical Engineering Services Letter, Geotechnical Drilling and Laboratory Services, Ahwahnee Hotel FEMA Evaluation, Yosemite, California”, 13 October.

REFERENCES (Cont.)

Kleinfelder (2009). “Geotechnical Investigation Report, Yosemite Valley Integrated Utility Project Phase 3B, Pre-Design and Design, Yosemite National Park”, 26 June.

Lienkaemper, J. J. (1992). “Map of Recently Active Traces of the Hayward Fault, Alameda and Contra Costa counties, California.” Miscellaneous Field Studies Map MF-2196.

Liu, L. and Dobry, R. (1997). “Seismic Response of Shallow Foundation on Liquefiable Sand.” Journal of Geotechnical and Geoenviromental Engineering, June.

Liu, N. and Mitchell, J.K. (2006). “Influence of Nonplastic Fines on Shear Wave Velocity – Based Assessment of Liquefaction.” Journal of Geotechnical and Geoenviromental Engineering, August.

McGuire, R.K. (2005). Personal communications.

McGuire, R. K. (1976). “FORTRAN Computer Program for Seismic Risk Analysis.” U.S. Geological Survey, Open-File Report 76-67.

McNutt, B. W. and Wilson, R. (1991), “Mono Lake Earthquake of October 23, 1990.” California Geology.

44(2):27-32.

Petersen, D., et al. (1996). “Probabilistic Seismic Hazard Assessment for the State of California,” USGS Open File Report 96-706.

Risk Engineering Inc. (2009). “EZFRISK computer program.” Version 7.34.

Rollins, K.M. and Seed, H.B. (1990). “Seismic Response of Shallow Foundation on Liquefiable Sand.”

Journal of Geotechnical Engineering, Vol. 116, No. 2, February.

Toppozada, T. R. and Borchardt G. (1998). “Re-Evaluation of the 1836 “Hayward Fault” and the 1838 San Andreas Fault earthquakes.” Bulletin of Seismological Society of America, 88(1), 140-159.

Townley, S. D. and Allen, M. W. (1939). “Descriptive Catalog of Earthquakes of the Pacific Coast of the United States 1769 to 1928.” Bulletin of the Seismological Society of America, 29(1).

Wells, D. L. and Coppersmith, K. J. (1994). “New Empirical Relationships among Magnitude, Rupture Length, Rupture Width, Rupture Area, and Surface Displacement.” Bulletin of the Seismological Society of America, 84(4), 974-1002.

Wesnousky, S. G. (1986). “Earthquakes, Quaternary Faults, and Seismic Hazards in California.” Journal of Geophysical Research, 91(1312).

Wieczorek, G. F., Snyder, J. B., Alger, C. S., and Isaacson, K. A. (1992). “Rockfalls in Yosemite Valley, California,” USGS Open File Report 92-387, 38 p., 2 appendixes, 4 plates, 1 disk.

Wieczorek, G. F. and Jaeger, S. (1996). “Triggering Mechanisms and Depositional Rates of Postglacial Slope-Movement Processes in the Yosemite Valley, California.” Geomorphology, 15:17-31.

Working Group on California Earthquake Probabilities (WGCEP) (2007). “The Uniform California Earthquake Rupture Forecast, Version 2.” Open File Report 2007-1437.

Working Group on California Earthquake Probabilities (WGCEP) (2003). “Summary of Earthquake Probabilities in the San Francisco Bay Region: 2002 to 2031.” Open File Report 03-214.

FIGURES

Project No. FigureDate

SITE LOCATION MAP

4994.0110/20/09 1

SITE

0 2,000 Feet

Approximate scale

1,000

Base map: Google Map, 2009

E dw ar d

S tr ee t

Treadwell&Rollo

AHWAHNEE HOTEL

Merced River

Rob in so n

Cre ek

M ono Lake

H artley S prings

H ilton C reek

R ound V alley

W hite M tns

F is h S lo ug h

O w ens Valley

Birch C reek

Damping Ratio = 5%

Note: Site Avg. Vs30 = 226 m/s

Project No. 4994.01 Figure 4Date 10/22/09

AHWAHNEE HOTEL

Yosemite National Park, California

RECOMMENDED SPECTRA

0.0

0.2

0.4

0.6

0.8

1.0

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0

PERIOD (seconds)

SP

EC

TR

A

L A

C C

EL

ER

A

TI

O N

(g 's

) BSE-2

BSE-1

APPENDIX A

Boring Logs

Project No. FigureDate A-4

CLASSIFICATION CHART

Major Divisions Symbols Typical Names

GW

GP

GM

GC

SW

SP

SM

SC

ML

CL

OL

MH

CH

OH

PTHighly Organic Soils

UNIFIED SOIL CLASSIFICATION SYSTEM

Well-graded gravels or gravel-sand mixtures, little or no fines

Poorly-graded gravels or gravel-sand mixtures, little or no fines

Silty gravels, gravel-sand-silt mixtures

Clayey gravels, gravel-sand-clay mixtures

Well-graded sands or gravelly sands, little or no fines

Poorly-graded sands or gravelly sands, little or no fines

Silty sands, sand-silt mixtures

Inorganic silts and clayey silts of low plasticity, sandy silts, gravelly silts

Inorganic clays of low to medium plasticity, gravelly clays, sandy clays, lean clays

Organic silts and organic silt-clays of low plasticity

Inorganic silts of high plasticity

Inorganic clays of high plasticity, fat clays

Organic silts and clays of high plasticity

Peat and other highly organic soils

Clayey sands, sand-clay mixtures

Range of Grain Sizes Grain Size in Millimeters U.S. Standard

Sieve Size Above 12"

12" to 3"

Classification

Boulders

Cobbles

Above 305

305 to 76.2

Silt and Clay Below No. 200 Below 0.075

GRAIN SIZE CHART

SAMPLER TYPE

C oa rs e-

G ra in ed

S oi ls (m or e th an h al f o f s oi l > n o.

si ev e si ze

Fi ne

-G ra in ed

S oi ls (m or e th an h al f o f s oi l no

. 2 s ie ve s iz e)

Gravels (More than half of coarse fraction >

no. 4 sieve size)

Sands (More than half of coarse fraction <

no. 4 sieve size)

Silts and Clays

LL = < 50

Silts and Clays

LL = > 50

Gravel coarse fine

3" to No. 4 3" to 3/4"

3/4" to No. 4

No. 4 to No. 200 No. 4 to No. 10 No. 10 to No. 40 No. 40 to No. 200

76.2 to 4.76

76.2 to 19.1

19.1 to 4.76

4.76 to 0.075

4.76 to 2.00

2.00 to 0.420

0.420 to 0.075

Sand coarse medium fine

C Core barrel

CA California split-barrel sampler with 2.5-inch outside diameter and a 1.93-inch inside diameter

D&M Dames & Moore piston sampler using 2.5-inch outside diameter, thin-walled tube

O Osterberg piston sampler using 3.0-inch outside diameter, thin-walled Shelby tube

PT Pitcher tube sampler using 3.0-inch outside diameter, thin-walled Shelby tube

S&H Sprague & Henwood split-barrel sampler with a 3.0-inch outside diameter and a 2.43-inch inside diameter

SPT Standard Penetration Test (SPT) split-barrel sampler with a 2.0-inch outside diameter and a 1.5-inch inside diameter

ST Shelby Tube (3.0-inch outside diameter, thin-walled tube) advanced with hydraulic pressure

SAMPLE DESIGNATIONS/SYMBOLS

Sample taken with Sprague & Henwood split-barrel sampler with a 3.0-inch outside diameter and a 2.43-inch inside diameter.

Darkened area indicates soil recovered

Classification sample taken with Standard Penetration Test sampler

Undisturbed sample taken with thin-walled tube

Disturbed sample, hand auger

Sampling attempted with no recovery

Core sample

Analytical laboratory sample

Sample taken with Direct Push sampler

Unstabilized groundwater level

Stabilized groundwater level

Treadwell&Rollo 4994.0109/08/09

APPENDIX B

Shear Wave Velocity Profile

0 50 100 150 200 250 300 350 400 450 500

VELOCITY (FEET/SECOND)

D

EP

TH

(M

ET

ER

S)

0 250 500 750 1000 1250 1500

VELOCITY (METERS/SECOND)

D

EP

TH

(F

EE

T)

Project No. FigureDateTreadwell&Rollo 4994.0110/22/09 B-1

SHEAR WAVE VELOCITY PROFILE

BORING TR-1

APPENDIX C

Laboratory Test Data

APPENDIX D

Site-Specific Ground Motions

D-1

APPENDIX D

SITE-SPECIFIC GROUND MOTIONS

This appendix presents the results of our site-specific ground motion studies for the proposed seismic strengthening at the Ahwahnee Hotel in Yosemite, California. The scope of our services was outlined in our proposals dated 14 July 2009 and 17 July 2009. We understand the provision of ASCE 41-06 will be used for seismic evaluation and design of the hotel. We also discussed our preliminary results in a meeting with Messrs. Loring Wylie and Christopher Poland of Degenkolb Engineers (Degenkolb), the project structural engineer, on 26 August 2009. During this meeting it was agreed that we would develop ground motion spectra for two levels of shaking. These were Basic Safety Earthquake 2 (BSE-2) and Basic Safety Earthquake 1 (BSE-1) both consistent with the definitions presented in ASCE 41-06 for each level of shaking. Furthermore, it was agreed that ground motions would be developed using the

Next Generation Attenuation (NGA) relationships for geometric mean and the comparison to the code spectra will be performed using the 2008 United States Geologic Survey (USGS) probabilistic maps.

D1.0 PROBABILISTIC AND DETERMINISTIC SEISMIC HAZARD ANALYSIS

The following sections present the details of our estimation of the level of ground shaking for this project.

We performed a Probabilistic Seismic Hazard Analysis (PSHA) and deterministic analysis to develop smooth, site-specific horizontal spectra for two levels of shaking, namely:

1. Basic Safety Earthquake 2 (BSE-2), which for this area in California corresponds to 2 percent probability of exceedance in 50 years (2,475 year return period);

2. Basic Safety Earthquake 1 (BSE-1) is defined as the lesser of 2/3 times BSE-2 or 10 percent probability of exceedance in 50 years (475 year return period).

D1.1 Probabilistic Seismic Hazard Analysis