GR 16-21 MT PRA GLAC 10(45) Final Geotech Report.pdf

Going-to-the-Sun Road & August 2021

Table of Contents 1-i

Geotechnical Report No. 16-21

Geotechnical Recommendations for the Replacement of McDonald Creek Bridge -

Rehabilitate Final 9.3 miles of the Going-to-the-Sun Road &

Replace Bridge Over McDonald Creek

GEOTECHNICAL ENGINEERING REPORT

MT PRA GLAC 10(45)

Glacier National Park, Montana

August 2021

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Geotechnical Recommendations for the Replacement of McDonald Creek Bridge August 2021

Table of Contents iii

TABLE OF CONTENTS

1 INTRODUCTION

1.1 Project Overview

1.2 Scope of Work

1.2.1 Reporting

1.2.2 Geotechnical Investigation

1.3 Project Datums

2 SITE CONDITIONS

2.1 Topography, Drainage, and Vegetation

2.2 Climate

3 SUBSURFACE CONDITIONS

3.1 Site Geology

3.2 Subsurface Investigation

3.2.1 Borings BH19-61 and BH19-62

3.2.2 Sampling and Testing

3.3 Subsurface Profile

3.3.1 West Abutment

3.3.2 East Abutment

3.4 Groundwater Observations

4 LABORATORY TESTING

4.1 Laboratory Testing

5 SEISMIC

5.1 Seismic Hazard

5.2 Liquefaction

5.3 Lateral Spreading

6 FOUNDATION RECOMMENDATIONS

6.1 Scour

6.2 Lateral Parameters

6.3 Axial Recommendations

6.3.1 Pipe Piles

6.3.2 Downdrag from Embankment Construction

6.3.3 Settlement

6.3.4 Global Stability

7 WING WALL RECOMMENDATIONS

8 MECHANICALLY STABALIZED EARTH WALL RECOMMENDATIONS

8.1 Construction Sequence

8.2 Global Stability

8.2.1 Excavation of Existing Embankment

GEOTECHNICAL ENGINEERING REPORT August 2021 iv Introduction

8.2.2 MSE Wall Construction

8.2.3 Final MSE Wall and Embankment

9 CONSTRUCTION CONSIDERATIONS

10 LIMITATIONS

11 REFERENCES

LIST OF FIGURES

Figure 1. Glacier National Park, project vicinity map

Figure 2. Mapped geology for the McDonald Creek Bridge Replacement project in Glacier National Park, Montana (Whipple, 1992)

Figure 3. 5% Damped Response Spectra – Site Class E

LIST OF TABLES

Table 1. Monthly Climate Summary for West Glacier, MT

Table 2. Summary of boring locations, elevations, and groundwater observations

Table 3. Soil laboratory testing summary

Table 4. Summary of Seismic Parameters (7% PE in 75 years) - General Procedure

Table 5. Factored axial Limit State Loads per abutment

Table 6. Recommended soil parameters for lateral analysis of piles

Table 7. Global stability C/D ratios for west and east abutments

Table 8. Global seismic stability C/D ratios for west and east abutments

Table 9. Liquefied shear strength weighting factors

Table 10. Global liquefied stability C/D ratios for the east abutment

Table 11. Recommended soil parameters for wing walls

Table 12. MSE wall parameters

Table 13. Initial temporary excavation static stability C/D ratios

Table 14. Static stability C/D ratios after MSE wall and north embankment construction

Table 15. Final MSE wall and embankment global stability C/D ratios

Introduction v

Appendices

APPENDIX A

Axial Foundation Capacity Charts

Foundation Global Stability Analysis Results MSE Wall and Embankment Global Stability Analysis Results

APPENDIX B

Boring Location Plan

APPENDIX C

Field Classifications for Soil and Rock Unified Soil Classification System Boring Logs BH19-61 and BH19-62

APPENDIX D

Laboratory Test Reports

Introduction 1

1 INTRODUCTION

1.1 PROJECT OVERVIEW

The McDonald Creek Bridge Replacement project comprises an approximately 0.3 mile section roadway approach and includes replacement of the existing bridge with a single span bridge approximately 213 feet long. The McDonald Creek Bridge Replacement project is part of the Rehabilitate Final 9.3 miles of the Going-to-the-Sun Road project, which also includes full depth reclamation with 25 curve enhancements and fiber utility upgrades. This report is specific to the McDonald Creek Bridge Replacement project, as shown in Figure 1.

The project area is located approximately 11 miles north-northeast of West Glacier, Flathead County, Montana west of the intersection of Going-to-the-Sun Road and Upper McDonald Creek Road in Glacier National Park. The coordinates of McDonald Creek Bridge are 45.6355° North, 113.8673° West.

The FHWA Bridge Inspection Report (2009) states the existing structure was constructed in 1965 and consists of a multi-span, timber girder bridge with an overall length of 201 feet. The existing bridge is founded on five bents consisting of vertical and battered timber piles. During the 2009 bridge inspection, a significant crack in a timber beam was observed. Because of this, the existing bridge was weight restricted to 12 tons.

The proposed structure will consist of a single-span, steel plate girder bridge with an overall length of 213-feet. The width will be increased to include a 14-foot-wide traveled width and two 5-foot-wide sidewalks. Because of the limited footprint available, the north half of the bridge will be constructed first to allow continued use of the existing bridge. A retaining wall will be required along the north side of the east embankment bridge approach. The proposed bridge will have an increased final grade height of approximately 8 feet.

2 Introduction

Figure 1. Glacier National Park, project vicinity map.

1.2 SCOPE OF WORK

1.2.1 Reporting

This report presents the geotechnical engineering recommendations for the replacement of McDonald Creek Bridge. This report is based on information provided by Robert Peccia & Associates (RPA) and subsurface investigations completed by Golder Associates. Sample testing was completed by FHWA WFLHD.

This will be a new structure having the following characteristics:

• Single span

• North half constructed first to maintain use of existing bridge

• Overall length of approximately 213 feet

• Overall width of 27 feet

• Single-lane, 14-foot travel way width

• Two, 5-foot wide sidewalks

• Parabolic steel plate girders

• Begin bridge station: 7+82.38

• End bridge station: 10+03.38

1.2.2 Geotechnical Investigation

The geotechnical exploration program was designed to provide information required to adequately characterize the subsurface conditions for the design and construction of the McDonald Creek Bridge and adjacent alignment. The location of the geotechnical borings are shown in Appendix B. The exploration was performed by Golder Associates October 22, 2019 through November 6, 2019 in conjunction with the geotechnical exploration for the Lake McDonald Utility Relocation project. The exploration program for the Replacement of McDonald Creek Bridge consisted of two borings/cores, BH19-61 and BH19-62, advanced to depths of 155 and 126 feet, respectively.

1.3 LABORATORY TESTING OF SAMPLES COLLECTED DURING DRILLING

OPERATIONS WAS PERFORMED TO EVALUATE INDEX PROPERTIES AND

STRENGTH CHARACTERISTICS OF THE SOIL. PROJECT DATUMS

The coordinate system is the Montana SPCS NAD83 1992. The vertical datum is based on the NAVD88 GEOID03. All elevations discussed in this report are approximate.

2 SITE CONDITIONS

2.1 TOPOGRAPHY, DRAINAGE, AND VEGETATION

McDonald Creek originates on the east side of Mount Geduhn within a valley of the Livingston Range. The creek generally flows southeast along the southwest valley of Flattop Mountain for

4 Site Conditions approximately 10 miles. McDonald Creek continues southwest within the valley between Heavens Peak to the west and Clements Mountain to the east for approximately 8 miles before discharging into Lake McDonald.

The topography of the greater project area generally consists of a northeast/southwest trending valley and ridge sequence. The drainage basin which encompasses McDonald Creek and recharges Lake McDonald is approximately 110 square miles. Information provided by RPA in the Preliminary Hydraulic Design Memorandum dated January 2021 stated a drainage basin peak flow for Q100 of 9,690 cfs.

Lake McDonald is located approximately 0.2 miles southwest of the existing bridge and subsequent proposed bridge. The lake is approximately 10 miles in length, 1.25 miles in width, and has a depth of 472 feet. The surface elevation of the lake was reported at approximately 3,159 feet.

At the location of the bridge, the bank full width of McDonald Creek is approximately 160 feet which increases to a width of up to 400 feet near the lake. Ordinary high water (OHW) was estimated by RPA for two scenarios: high lake elevation and low lake elevation. OHW was estimated at 3,162 feet and 3,160 feet, which is approximately 10 feet below the proposed bridge low chord elevations. The west bank is 16 feet in height with a slope of about 1H:1V (horizontal:vertical) and observable erosion of the bank material. The east bank is also about 16 feet in height with a shallower slope of 2H:1V and minimal erosion.

The terrain north of the west bridge approach is relatively level and is vegetated with grasses and deciduous and coniferous trees up to approximately 2 feet in diameter. The terrain south of the west bridge approach slopes downward 6 feet at 3H:1V. At the bottom of the shallow slope, there is a clearing vegetated with grasses that was once the roadway for a previous bridge that was washed out during a flood event in 1964. The clearing continues to parallel the existing roadway from the river for approximately 150 feet, where it intersects with the existing roadway.

The terrain on both sides of the east bridge approach embankment slope downward approximately 10 feet at 2H:1V. The area is vegetated with grasses, brush, and deciduous and coniferous trees up to 2 feet in diameter. At the bottom of the slope on the north side of the embankment, the ground surface appears to be the natural surface elevation prior to the construction of the existing embankment.

2.2 CLIMATE

Table 1 summarizes average temperatures, precipitation, and snowfall measurements from the West Glacier weather station (WBAN: 248809), located approximately 11 miles southwest of the site.

Table 1. Monthly Climate Summary for West Glacier, MT.

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec Annual

Average Max.

Temperature (°F)

30.5 35.4 43.5 54.1 64.4 71.9 80.0 79.0 67.5 52.7 36.7 29.5 53.9

Average Min.

Temperature

(°F)

18.0 19.4 24.7 30.7 37.9 44.1 48.4 46.9 39.4 32.1 24.8 18.4 32.2

Extreme Max.

Temperature (°F)

55 58 66 83 90 93 99 99 95 79 65 52 99

Extreme Min.

Temperature

(°F) -35 -32 -30 3 13 24 31 26 18 -3 -29 -36 -36

Average Total

Precipitation (in.)

2.99 2.11 2.00 1.95 2.65 3.36 1.76 1.40 2.00 2.49 3.19 3.13 29.03

Average Total

Snowfall (in.)

36.1 21.6 13.9 3.2 0.5 0.0 0.0 0.0 0.1 2.0 16.4 34.8 131.4

Notes: Period of record 1948 to 2016.

Extreme max and min temperature, and average total snowfall are period of record data.

Average max and min temperature, and average total precipitation are 1981 to 2010 data.

3 SUBSURFACE CONDITIONS

3.1 SITE GEOLOGY

According to the Geologic Map of Glacier National Park, Montana (Whipple, 1992), the project area is mapped as “Qg” and relatively smaller areas are mapped as “Qal,” specifically along the shores of Upper McDonald Creek as shown in Figure 2.

The “Qg” unit is defined as Till (Holocene and upper Pleistocene). It is described as unsorted, subrounded to subangular, boulder rubble with small amounts of sand, silt, and clay. The unit can be up to 200 feet thick. The age range is reported as up to 129,000 years old.

The “Qal” unit is defined as Alluvium (Holocene and upper Pleistocene). It is described as rounded to subrounded sand and gravel deposits locally containing thin lenses of silt. This includes channel deposits in modern floodplains. The typical thickness of this unit is reported to be up to 33 feet thick. The age range is also reported as up to 129,000 years old.

The ”Yapp” unit is defined as the Appekunny and Prichard Formations, undivided and is mapped approximately 750 feet upstream of the project site along Upper McDonald Creek. It is interpreted the bedrock in the borings encountered the lower part of this mapped unit, which is the Prichard Formation described as thin, even, parallel laminae of rusty-weathering, blackish-gray argillite and light gray siltite. The age range is reported as 1.3 to 1.7 billion years old.

6 Subsurface Conditions

3.2 SUBSURFACE INVESTIGATION

The subsurface investigation for this project consisted of two drilled borings that were completed in conjunction with the Lake McDonald Utility Relocation project. The two borings, BH19-61 and BH19-62, were completed October 22, 2019 through November 6, 2019. The boring locations are summarized in Table 2 and shown on the Boring Locations Map in Appendix B. The boring logs are presented in Appendix C.

Project Site

BH19-62

BH19-61

Figure 2. Mapped geology for the McDonald Creek Bridge Replacement project in Glacier National Park, Montana (Whipple, 1992).

Table 2. Summary of boring locations, elevations, and groundwater observations.

Boring No.

General Location Latitude Longitude

Ground Surface

Elevation (feet)

Boring Depth (feet)

Water Level Observations

Depth (feet)

Elevation (feet)

BH19-61 West Abutment 48.6355° -113.8673° 3,175 154.9 18.2 3,156.8

BH19-62 East Abutment 48.6352° -113.8665° 3,172 126.0 16.1 3,155.9

3.2.1 Borings BH19-61 and BH19-62

The borings were completed by Haz-Tech Drilling of Meridian, Idaho in two phases. Boring BH19- 61 was completed October 22-24, 2019, and BH19-62 was completed November 5-6, 2019. The borings were performed with a CME-75 truck mounted drill rig. Borings were advanced with 8-inch diameter hollow stem auger and 4-inch diameter casing advancer in overburden materials. HQ wireline coring was used to advance the holes and retrieve samples in bedrock. Groundwater readings were taken 15 minutes after the completion of drilling. Borings were backfilled with soil cuttings and bentonite chips following completion.

The borings were logged in the field by a geotechnical engineer from Golder Associates in accordance with the Federal Lands Highway (FLH) Soil and Rock Characterization Guidelines (2015).

3.2.2 Sampling and Testing

Standard Penetration Tests (SPTs) were completed at 2.5-foot to 5-foot intervals in overburden material. The Standard Penetration Test (SPT) consists of using a 140-pound hammer with a 30-inch drop to drive a two inch outside diameter split-barrel sampler 18 inches below the bottom of the bore hole and recording the number of blows for each six-inch increment of penetration. The total number of blows required to drive the sampler the second and third 6-inch intervals is known as the N-value (blows per foot).

Samples collected during all phases were shipped to WFLHD in Vancouver, WA. The samples and draft borings logs were reviewed by the WFLHD geotechnical engineer and engineering geologists assigned to the project. Selected samples of soil were submitted to the WFLHD materials laboratory for testing.

3.3 SUBSURFACE PROFILE

The borings encountered material interpreted as fill and native alluvial deposits overlying bedrock.

8 Subsurface Conditions

3.3.1 West Abutment

Boring BH19-61 was completed near the west bridge abutment. The subsurface observations for this boring generally consisted of loose to medium dense, gray and brown sand with varying silt and gravel content, interpreted as alluvial deposits, beginning at the ground surface. Underlying the alluvium at Elevation 3,042 feet (depth of 133 feet), the boring encountered gray, very stiff lean clay. At Elevation 3,030 feet, the boring encountered gray, medium strong siltite rock interpreted as Prichard Formation bedrock. The boring was terminated within this rock at Elevation 3,020 feet (depth of 155 feet). The observed overburden and rock are consistent with mapped geology.

3.3.2 East Abutment

Boring BH19-62 was completed near the east bridge abutment. Underlying approximately 0.5 feet of asphalt concrete, the boring encountered medium dense, brown sand with gravel, interpreted as fill. Underlying the fill at Elevation 3,160 (depth of 12 feet), the boring encountered a 2.5-foot-thick layer of firm, dark brown organic silt, interpreted as lacustrine deposits.

At Elevation 3,157.5 feet (depth of 14.5 feet), the boring encountered loose to medium dense, brown sand with varying silt and gravel content, interpreted as alluvial deposits. Underlying the sand at Elevation 3,112.5 (depth of 59.5 feet), the boring encountered a 5-foot-thick layer of medium dense, brown gravel with sand overlying a 5-foot-thick layer of firm silt. At Elevation 3,104 feet (depth of 68 feet), the boring encountered medium dense silty sand with laminated layers of trace organics up to ¾ -inch thick, which consisted of woody debris. Underlying the silty sand at Elevation 3,078 feet (depth of 94 feet), the boring encountered a 9-foot-thick layer of very soft to soft, light gray lean clay with moderate plasticity.

The boring encountered a 10-foot thick layer of very dense, gray silty sand, interpreted as till at Elevation 3,066.5 feet (depth of 105.5 feet). Underlying the till, the boring encountered strong, gray siltite rock, interpreted as Prichard Formation bedrock. The boring was terminated within this rock unit at Elevation 3,046 feet (depth of 126 feet). The observed overburden and rock are consistent with mapped geology.

3.4 GROUNDWATER OBSERVATIONS

Water level observations made during the subsurface investigation are summarized in Table 2.

Groundwater was observed in the borings at approximately Elevation 3,156 feet, which was similar to the water elevation of McDonald Creek at the time of drilling.

4 LABORATORY TESTING

4.1 LABORATORY TESTING

Selected split-barrel (SPT) soil samples were submitted to the WFLHD Materials Laboratory for routine laboratory testing. Laboratory test results are discussed below, and test reports are contained in Appendix D.

The following laboratory tests were performed on the selected soil samples:

• Soil Classification (AASHTO M145)

• Sieve Analysis (AASHTO T11/T27)

• Atterberg Limits (AASHTO T89/T90)

• Direct Shear (AASHTO T236)

• Moisture Content (AASHTO T265)

• Organic Content (AASHTO T267)

• Resistivity (AASHTO T288)

• pH of Soil (AASHTO T289)

The results of the soil tests were used to classify the samples according the Unified Soil Classification System (USCS). The USCS classification is reported on the boring logs. A summary of the soil testing results is presented in Table 3. The soil types displayed in Table 3 varied but primarily consisted of silty, clayey Sand, silty, clayey Gravel, or lean Clay with sand.

Table 3. Soil laboratory testing summary.

Test Tests Completed Range of Values

AL – Plastic Limit 8 NP to 37%

AL – Liquid Limit 8 NP to 45%

AL– Plasticity Index 8 NP to 16%

Direct Shear 2 46.9° to 49.4°

Moisture Content 17 15.0% to 68.4%

Organic Content 3 2.2% to 10.9%

Resistivity 1 4,400 ohm-cm pH 3 5.3 to 6.4

10 Seismic

5 SEISMIC

5.1 SEISMIC HAZARD

The General Procedure outlined in Section 3.10.2 of the AASHTO LRFD Bridge Design Specifications, 9th Edition (2020) was followed to characterize the seismic hazard. The General Procedure uses mapped gridded values of peak ground acceleration, 0.2 second spectral acceleration, and 1.0 second spectral acceleration to develop the 5-percent-damped-design response spectrum chart.

These procedures use a design earthquake with a return period of 975 years, or a 7 percent probability of exceedance (PE) in 75 years. The USGS Seismic Design Reports Application was used to determine the peak ground acceleration and spectral accelerations (USGS, 2017). Site Class E was selected at the abutment locations based on the N60 standard penetration resistance of the soil in general accordance with Section 3.10.3, Method B (AASHTO, 2020). Site factors were selected from Table 3.10.3.2-1 through 3.10.3.2-3 in AASHTO (2020) and multiplied by the mapped peak ground accelerations and spectral accelerations in order to determine the modified peak ground acceleration and spectral accelerations. The results of the hazard analysis are presented in Table 4 and Figure 3.

Table 4. Summary of Seismic Parameters (7% PE in 75 years) - General Procedure.

Site Latitude 45.6355° Site Longitude -113.8673°

Site Class at Project Site by Soil Type and Penetration Resistance in Top 100 feet of Soil Profile) E

Horizontal Peak Ground Acceleration on Site Class B Rock (PGA) 0.091 g

Horizontal Response Spectral Acceleration at a Period of 0.2 sec on Site Class B Rock (Ss) 0.214 g

Horizontal Response Spectral Acceleration at a Period of 1.0 sec on Site Class B Rock (S1) 0.079 g

Mapped Peak Ground Acceleration Coefficient, FPGA 2.500 Mapped Short-Period Spectral Acceleration Coefficient, Fa 2.500 Mapped 1-sec Period Spectral Acceleration Coefficient, Fv 3.500

Scaled Peak Ground Acceleration for Project Site, Site Class E (As) 0.228 g

Scaled Horizontal Response Spectral Acceleration at a Period of 0.2 sec for Project Site, Site Class E (SDS) 0.535 g

Scaled Horizontal Response Spectral Acceleration at a Period of 1.0 sec for Project Site, Site Class E (SD1) 0.277 g

AASHTO Seismic Zone 2

Figure 3. 5% Damped Response Spectra – Site Class E.

5.2 LIQUEFACTION

Soil liquefaction refers to the condition by which vibration or shaking of the ground, usually from earthquake forces, results in the collapse of the soil structure, the development of excess pore pressures in saturated soils, and a subsequent reduction in soil shear strength. Evaluation of liquefaction potential is dependent on numerous parameters, including soil type, grain-size distribution, soil density, depth to groundwater, in-situ static ground stresses, earthquake magnitude, peak ground acceleration (PGA), and associated earthquake-induced ground stresses. In general, soils that are susceptible to liquefaction include very loose to medium dense, clean to silty sands and very soft to stiff non-plastic silts that are below the water table.

Typically, liquefaction potential at a site is completed using simplified methods, e.g. Youd and Idriss (2001) and Idriss and Boulanger (2014). These methods compare the cyclic resistance ratio (CRR) of a soil layer (the cyclic shear stress required to trigger liquefaction) to the cyclic stress ratio (CSR) induced by an earthquake. The factor of safety is estimated by dividing the CRR by the CSR.

Using the simplified methods above, the liquefaction analysis indicated that there is liquefaction potential at the east abutment. However, the design ground motion (Mw, mode = 5.1) is outside of the case histories used to develop the simplified methods. Therefore, the simplified methods are not considered appropriated for the project seismic design event. Although a fully liquefied condition is not expected, a design event could induce elevated pore pressures which may result in up to 2-inches total and 1-inch of differential settlement.

0.00 0.10 0.20 0.30 0.40 0.50 0.60 0.70 0.80 0.90 1.00 1.10 1.20

0.0 0.2 0.4 0.6 0.8 1.0 1.2 1.4 1.6 1.8 2.0

Sp ec tr al

A cc el er at io n, S a (g

Period, T (s)

12 Foundation Recommendations

5.3 LATERAL SPREADING

Lateral spreading related to seismic activity typically involves lateral displacement of large, surficial blocks of non-liquefied soil when a layer of underlying soil loses strength during seismic shaking. Lateral spreading usually develops in areas where sloping ground or large grade changes are present. Based on the planned improvements, site topography, and liquefaction potential explained above, the risk of lateral spreading impacting the site is low.

6 FOUNDATION RECOMMENDATIONS

Foundation recommendations are discussed below. The bridge abutments are expected to be founded on 30-inch diameter steel pipe piles. RPA provided final loads for the proposed bridge via email dated August 18, 2021. Factored axial Limit State loads per abutment representing the half construction condition and fully constructed condition are presented in Table 5

Table 5. Factored axial Limit State Loads per abutment.

Limit State Half Width Constructed

Condition Abutment Load (kips)

Full Width Constructed Condition

Abutment Load (kips) Axial Service I 611 1,239

Axial Strength I 807 1,673

Axial Extreme I 568 1,130

6.1 SCOUR

RPA provided scour recommendations in their Final Hydraulic Design Memorandum dated April 2021. The total scour elevation for the 200-year event flow recommended in the memo was approximately Elevation 3,150 feet and 3,145 feet in reference to project datum for the west and east abutments, respectively. It is proposed that riprap will be installed along the bank as a risk reduction method against erosion.

6.2 LATERAL PARAMETERS

Table 6 presents the soil parameters recommended to analyze lateral loading and foundation capacity with software such as LPile. The values in Table 6 do not include factors of safety or load factors. A resistance factor of 1.0 is specified in AASHTO (2020) for lateral geotechnical resistance of single or group shafts. Because the east abutment may experience minimal liquefaction, it is recommended to apply a P-multiplier, Pm, of 0.1 from Elevation 3,147 feet to Elevation 3,140 feet for Extreme Limit State I. The reduction in lateral capacity due to liquefaction should not be combined with scour.

For the scour load case, lateral support should be ignored to Elevation 3,145 feet and 3,150 feet for the west and east abutments, respectively.

For lateral pile or shaft loading, reductions in the soil response p-y curves are necessary if the piles are spaced close enough to influence the other piles in the group. The values of P shall be multiplied by Pm to account for group effects. The values of Pm presented below should be used, if applicable.

Longitudinal Direction

For pile loading in the longitudinal direction, it is necessary to reduce the P values in the soil response p-y curves if the center-to-center spacing of the piles is less than five equivalent pile diameters. The p-y multiplier can be determined from the equation below:

Pm = 0.1(S) + 0.5

Where:

Pm = p-y multiplier

S = center-to-center pile spacing, expressed in number of equivalent pile diameters

Transverse Direction

For pile/shaft loading in the transverse direction, it is necessary to reduce the P values in the soil response p-y curves if the center to center spacing of the piles/shafts is less than 5 equivalent pile/shaft diameters. The p-y multiplier for the first pile/shaft can be determined from the equation below:

Pm = 0.1(S) + 0.5

The second pile will be influenced by the “shadow” of the first pile if the pile spacing is less than

5.7 equivalent pile diameters. The p-y multiplier for the second pile/shaft can be determined from the equation below:

Pm = 0.225(S) - 0.275

The third and subsequent piles/shafts will be influenced by the “shadow” of the piles/shafts in front if the pile/shaft spacing is less than 6.5 equivalent pile/shaft diameters. The p-y multiplier for the third and subsequent piles/shafts can be determined from the equation below:

Pm = 0.2(S) - 0.3

14 Foundation Recommendations

Table 6. Recommended soil parameters for lateral analysis of piles.

Elev. (ft) LPile Material γ (pcf) γ’ (pcf) φ’ (deg) k (lb/in3)

West Foundation (ground surface assumed = Elev. 3,175 feet, use effective unit weight below Elev. 3,162 feet)

>3,162 Sand (Reese) 120 - 34 90 3,162-3,117 Sand (Reese) - 58 34 60 3,117-3,038 Sand (Reese) - 60 36 75

East Foundation (ground surface assumed = Elev. 3,172 feet, use effective unit weight below Elev. 3,162 feet)

>3,162 Sand (Reese) 120 - 34 100 3,157-3,111 Sand (Reese) - 58 36 60 3,111-3,046 Sand (Reese) - 60 36 60

NOTES:

1. γ = moist unit weight.

2. γ’ = effective unit weight.

3. φ’ = effective internal angle of friction.

4. k is the constant giving the variation of soil modulus with depth in the equation Es = kz where Es = soil modulus, and z = depth below the ground surface.

6.3 AXIAL RECOMMENDATIONS

6.3.1 Pipe Piles

It is anticipated that traditional deep foundation installation equipment will be able to access the east side abutment. Factored resistances were developed using the following resistance factors in accordance with AASHTO LRFD BDS (2020):

• Service I Limit State – 1.00

• Strength I Limit State – 0.50

• Extreme I Limit State – 1.00

The above resistance factors were used to develop estimated piled lengths based on calculated capacities. Based on the loads provided and subsurface conditions observed, a total of 4, 30-inch diameter, 1/2-inch-thick sidewall steel pipe piles are recommended to be installed to a minimum tip Elevation of 3,125 feet. To achieve capacity, the estimated tip Elevation is 3,095 feet. As stated previously, minimal liquefaction is possible at the east abutment location. To account for this, a reduction in geotechnical pile capacity and associated downdrag loads have been applied to Elevation 3,140 feet for the Extreme I limit state. Factored individual pipe pile axial capacity charts presented in Appendix A.

Dynamic load testing will be required for one pile per abutment. Therefore, a resistance factor of

0.65 was applied to develop the required nominal driving resistance of 785 kips to be achieved during installation. The nominal driving resistance assumes full scour depth resulting in driving resistances greater than the nominal resistances.

6.3.2 Downdrag from Embankment Construction

Based on the location and depth of the foundation elements relative to the proposed embankments, downdrag loads resulting from embankment settlement are not anticipated.

6.3.3 Settlement

An evaluation of settlement for foundation elements has been performed. Settlement is anticipated to occur immediately (during construction) and is estimated to be less than 0.5 inches.

6.3.4 Global Stability

Global slope stability analyses were performed for the west and east abutments. The slopes were analyzed for static, seismic (pseudo-static), and liquefied stability using Slide2 (Rocscience, Inc.

2020). Three force/moment equilibrium methods were used: Bishop simplified, Spencer, and GLE/Morgenstern-Price. Soil properties used in the analyses were informed by the subsurface exploration boring logs and associated laboratory results. The geometry was developed from the preliminary plan set dated January 2021. Riprap geometry was developed from preliminary drawings dated March 2021. If the riprap geometry changes as the design progresses, global stability may have to be reevaluated. Results of the global stability analyses are discussed below and presented in Appendix A.

Static Stability

Under static conditions a minimum capacity to demand (C/D) ratio of 1.5 is required for the slope to be considered stable. A summary of the C/D ratios for the west and east abutments are displayed in Table 7.

Table 7. Global stability C/D ratios for west and east abutments.

Global Static Stability Method West Abutment East Abutment

Bishop Simplified 1.50 1.67

Spencer 1.51 1.70

GLE / Morgenstern-Price 1.51 1.72

These results were determined without using the shear resistance of the piles as a stabilizing force, and therefore these results are considered conservative. The analysis shows that the stability at both abutments is acceptable.

16 Foundation Recommendations

Seismic Stability

Pseudo-static slope stability analyses were performed at the west and east abutments using the estimated drained strength of the foundation soils and a horizontal seismic coefficient of 0.11 g, which is half of the calculated surface peak ground acceleration based on values presented in Table 4. This analysis models conditions during an earthquake before loss in soil strength occurs.

Slope displacement is expected to be less than two inches if the C/D ratio is greater than 1.1. A summary of the C/D ratios for the west and east abutments are displayed Table 8.

Table 8. Global seismic stability C/D ratios for west and east abutments.

Global Seismic Stability Method West Abutment East Abutment

Bishop Simplified 1.14 1.27

Spencer 1.15 1.31

GLE / Morgenstern-Price 1.15 1.33

These results were determined without using the shear resistance of the piles as a stabilizing force; therefore, these results are considered conservative. The analysis shows that the stability at both abutments is acceptable.

Liquefied Stability

As discussed in Section 5.2 of this report, minimal liquefaction may occur as a result of the design seismic event at the east abutment. A liquefied stability analysis was completed using liquefied shear strength soil properties. The liquefied shear strength properties were estimated using weighted values from simplified procedures proposed by Olson and Stark (2002), Idriss and Boulanger (2014), Kramer and Wang Hybrid (2008), and Kramer and Wang (2015). Weighting factors for each method are presented in Table 9.

Table 9. Liquefied shear strength weighting factors.

Procedure Weighting Factor

Olson and Stark 0.2

Idriss and Boulanger 0.2

Kramer and Wang, Hybrid 0.2

Kramer and Wang 0.4

As shown in these methods and presented by Stark et. al (2020), the liquefied shear strength is a function of overburden/consolidation stress for most cases. Therefore, a reduced liquefied friction angle was used in the Slide2 (Rocscience, Inc. 2020) model as opposed to applying a liquefied shear strength that remains constant throughout the depth of the liquefied layer. It is assumed that the design seismic event is relatively short and soil strength loss will not occur during shaking. Therefore, the liquefied stability analysis is decoupled form horizonal seismic accelerations. A minimum C/D ratio of 1.1 is considered stable for the liquefied scenario. A summary of the liquefied stability C/D ratios for the east abutment are displayed Table 10.

Table 10. Global liquefied stability C/D ratios for the east abutment.

Global Liquefied Stability Method East Abutment

Bishop Simplified 1.40

Spencer 1.45

GLE / Morgenstern-Price 1.41

These results were determined without using the shear resistance of the piles as a stabilizing force; therefore, these results are considered conservative. The analysis shows that the liquefied global stability at the east abutment is acceptable.

7 WING WALL RECOMMENDATIONS

Select borrow should be used behind wing walls. Geocomposite sheet drains connected to drain pipe and weep holes should be installed behind walls to ensure adequate drainage. The soil parameters listed in Table 11 should be used to estimate the lateral earth pressures exerted on the abutment walls and wing walls. The parameters listed in Table 11 assume a vertical wall with level ground at the top of the wall. Passive pressure should be neglected from the surface to a depth of 2 feet.

Table 11. Recommended soil parameters for wing walls.

Soil Parameter Design Value Unit weight 120 pcf

Angle of internal friction 36 degrees Active earth pressure coefficient, Ka 0.24

Active equivalent fluid pressure 29 psf/ft At-rest earth pressure coefficient, Ko 0.42

At-rest equivalent fluid pressure 50 psf/ft Passive earth pressure coefficient, Kp 10.00

Passive equivalent fluid pressure 1,000 psf/ft Wall-backfill interface friction angle, δ 27 degrees

8 MECHANICALLY STABALIZED EARTH WALL

RECOMMENDATIONS

Because of the limited footprint of the proposed structure, the east approach embankment will require a retaining wall from STA 6+80 to STA 7+68 along the north side. A mechanically stabilized earth (MSE) wall constructed using a welded wire face is recommended. The facing

18 mechanically stabalized earth wall recommendations should consist of dark rock with non-galvanized steel welded wire. Due to Park aesthetic requirements, other facing types will not be allowed. Global stability analyses representing different stages of construction were completed using Slide2 (Rocscience, Inc. 2020). The geometry was developed from the preliminary plan set dated January 2021. Global stability analyses are discussed below and included in Appendix A.

MSE wall layout values are listed in Table 12. Organic silt was observed in boring BH19-62 from Elevation 3,160 feet to Elevation 3,157 feet. It is recommended to remove the full depth of organic silt from 10 feet in front of the wall to 20 feet behind the face of the wall. It is assumed the MSE wall will be contractor designed.

Table 12. MSE wall parameters.

Bottom of Wall Elevation 3,157 ft Facing Non-galvanized Welded Wire

Minimum Reinforcement Length 10 ft Maximum Vertical Reinforcement spacing, Sv 2 ft

Minimum Backfill Material Internal Friction 34 degrees

8.1 CONSTRUCTION SEQUENCE

During construction of the north half of the bridge, the existing bridge will be open to traffic. Based on this, it is assumed that the sequence of construction for the MSE wall will be as follows:

• closure of the north portion of the existing embankment;

• excavation of the existing embankment fill material from the north side of the embankment and underlying organic silt to approximate Elevation 3,157 feet, actual elevation will be determined by depth of organic silt encountered during construction;

• construction of the MSE wall;

• closure of the south portion of the existing embankment; and

• construction of the south potion of the proposed embankment.

If the construction sequence is different than anticipated in this report, MSE wall global stability analyses may require reevaluation.

8.2 GLOBAL STABILITY

Global slope stability analyses were performed for the west and east abutments. The slopes and MSE wall were analyzed for static, seismic (pseudo-static), and liquefied stability using Slide2 (Rocscience, Inc. 2020). Three force/moment equilibrium methods were used: Bishop simplified, Spencer, and GLE/Morgenstern-Price. Soil properties used in the analyses were informed by the subsurface exploration boring logs and associated laboratory results. The geometry was developed from the preliminary plan set dated January 2021. Results of the global stability analyses are discussed below and presented in Appendix A.

8.2.1 Excavation of Existing Embankment

The north portion of the existing embankment should be excavated at a slope of 1V:1.5H. If soil conditions allow, the temporary excavation slope may be steepened to 1V:1.3H. Depending on site geometry, temporary walls may be required. It is anticipated that temporary walls will be less than 5 feet in height. If a temporary wall is required, the contractor must submit stability analysis results to the CO for approval. All excavations should be performed in accordance with OSHA regulations.

The organic silt observed in boring BH19-61 should be removed from 10 feet in front of and 20 feet behind the proposed MSE wall face. This is considered a temporary condition; and therefore, seismic and liquefied stability analyses are not required. A traffic surcharge was applied to the existing embankment because it will be open to traffic. A minimum global static stability C/D ratio of 1.1 is considered stable. Results for the temporary initial static stability analyses are presented in Table 13.

Table 13. Initial temporary excavation static stability C/D ratios.

Static Stability Method Initial Excavation

Bishop Simplified 1.18

Spencer 1.17

GLE / Morgenstern-Price 1.17

The analysis shows that the global stability for the initial temporary excavation is acceptable.

8.2.2 MSE Wall Construction

This scenario includes construction of the MSE wall and north portion of the proposed embankment. This is also considered a temporary condition; and therefore, seismic and liquefied stability analyses are not required. A traffic surcharge was applied to the north embankment and removed from the south embankment. A minimum global static stability C/D ratio of 1.1 is considered stable. Results for the temporary stability analyses of the slope after MSE wall construction are presented in Table 14.

20 Construction Considerations

Table 14. Static stability C/D ratios after MSE wall and north embankment construction.

Static Stability Method MSE Wall and North Embankment

Bishop Simplified 1.58

Spencer 1.56

GLE / Morgenstern-Price 1.58

The analysis shows that the global stability for the temporary slope after MSE wall construction is acceptable.

8.2.3 Final MSE Wall and Embankment

This scenario includes the finished construction of the MSE wall and embankment. The MSE wall and embankment were analyzed for static, seismic (pseudo-static), and liquefied stability following the same procedures stability requirements discussed in Section 6.3.5 Global Stability of this report. The liquefied scenario stability results are the same as the static stability results because the depth to the liquefied layer is much deeper than the static critical slip surface. Results of the static, seismic, and liquefied analyses are presented in Table 15. Details are included in Appendix A.

Table 15. Final MSE wall and embankment global stability C/D ratios.

Global Stability Method Static Seismic Liquefied

MSE

Wall

South Slope

MSE

Wall

South Slope

MSE

Wall

South Slope

Bishop Simplified 1.60 1.66 1.25 1.29 1.60 1.66

Spencer 1.59 1.66 1.24 1.29 1.59 1.66

GLE / Morgenstern-Price 1.59 1.66 1.24 1.29 1.59 1.66

The analyses show that the global stability for the MSE wall and embankment is acceptable.

9 CONSTRUCTION CONSIDERATIONS

9.1 PIPE PILES

Based on the subsurface investigation, the contractor should be prepared to remove possible obstructions within the initial 15 feet of driving. Possible obstructions noted in the boring logs include cobbles and boulders up to 14 inches in diameter. The contractor should be prepared to encounter riprap boulder obstructions up to 3 feet in diameter.

The piles are required to be driven using nonimpact hammers. Impact hammers will only be used for dynamic load testing to confirm capacity. Testing must be successfully completed on one pile per abutment.

Pile installation sequence:

1. Advance test pile with nonimpact hammer to minimum pile tip elevation.

2. Allow test pile to set up overnight.

3. Complete dynamic load test with impact hammer.

4. Provide testing results to CO within 24 hours.

5. 24 hours after receiving test results, CO will provide direction based on the following scenarios:

o Pile has achieved capacity. Install remaining piles for that abutment with nonimpact hammer to the same pile tip elevation as the test pile.

o Pile has not achieved capacity. Use nonimpact hammer to advance pile to estimated pile tip elevation, or a different elevation as directed by the CO. Repeat sequence from Step 2.

Piles must be protected from damage. If all piles are installed at the same time, the Bridge Section Phase 2 piles should be protected by temporarily backfilling voids inside of the pile and covering the pile tops with at least 2 feet of soil. Locations where traffic is expected to pass over the piles, a steel sheet should be placed to bridge the piles.

9.2 EMBANKMENT

Embankments should be constructed with backfill material and 1V:2H slopes. It is anticipated that excavated material from the existing east embankment may be suitable as backfill material. If the excavated soil is to be used, it should be free of organics and deleterious material.

9.3 TEMPORARY WALLS

If temporary walls are required during construction, the walls must be removed and must not be left in place and buried. If temporary MSE walls are used, the facing and exposed reinforcement must be horizontally cut at each layer prior to continuing construction of the embankment.

9.4 WET WEATHER EARTHWORK

Based on the subsurface investigation, the existing soils at the project site generally contain a low fines content; however, the soils encountered may still be difficult to work with and susceptible to disturbance when wet. During embankment excavation of the organic silt layer for the MSE wall, standard dewatering methods may be required to adequately prepare the wall subgrade.

10 LIMITATIONS

This report has been prepared to present final geotechnical engineering recommendations for this project. It should not be used, in part or in whole for other purposes without contacting the WFL Geotechnical Section for a review of the applicability of such reuse.

22 Limitations

Subsurface exploration and associated laboratory testing describes subsurface conditions only at the sites of subsurface exploration and at the intervals where samples are collected. These data are interpreted by members of the WFL Geotechnical Section, who then render a professional geotechnical opinion regarding the general subsurface conditions. The distribution, continuity, thickness, and characteristics of identified (and unidentified) subsurface materials may vary considerably from that indicated by the subsurface data. In addition, the observed water levels and/or conditions indicated on the test boring logs are recorded at the time of the exploration. The water levels and/or conditions may vary considerably, with time, according to the climate, rainfall, and other factors and are otherwise dependent on the duration of, and methods used in, and interpretations made from the subsurface exploration program.

This report should be made available to prospective contractors for their information, or as factual data only, and not as a warranty of ground conditions.

Questions regarding this report should be addressed to Sean Hayter, WFLHD Geotechnical Engineer at (202) 934-0377 or Sean.Hayter@dot.gov.

GEOTECHNICAL REPORT NO. 16-21

Geotechnical Recommendations for the Replacement of McDonald Creek Bridge - Rehabilitate Final 9.3 miles of the Going-to-the-Sun Road & Replace Bridge Over McDonald Creek

MT PRA GLAC 10(45)

Glacier National Park, Montana August 2021

Prepared by: ________________________________________________ Sean Hayter, Geotechnical Engineer

Reviewed by: ________________________________________________ Evan Garich, Geotechnical Engineer

Approved by: ________________________________________________ Douglas Anderson, Geotechnical Functional Manager

24 References

11 REFERENCES

AASHTO (2020), AASHTO LRFD Bridge Design Specifications, 9th Edition, American Association of State Highway and Transportation Officials, Washington, D.C.

AASHTO (2020), Standard Specifications for Transportation Materials and Methods of Sampling and Testing, American Association of State Highway and Transportation Officials, Washington D.C.

ASTM International (2020), Annual Book of ASTM Standards, Section Four, Construction, Volume

4.08 and 4.09 Soil and Rock.

Boulanger, R. W., and Idriss, I. M. (2014), CPT and SPT Based Liquefaction Triggering Procedures, Department of Civil and Environmental Engineering, University of California, Davis, California.

FLH (2017), Federal Lands Highway Soil and Rock Characterization Guidelines, Federal Highway Administration, December 2017.

Kramer, S. L., and C. H. Wang (2008), Evaluation of Liquefaction Hazards in Washington State, Washington State Department of Transportation Research Report, WA-RD 668.1.

Kramer, S. L., and C. H. Wang (2015), Empirical Model for Estimation of the Residual Strength of Liquefied Soil, Journal of Geotechnical and Geoenvironmental Engineering, ASCE 141 (9):

04015038.

Olson, S. M., and T. D. Stark (2002), Liquefied Strength Ratio from Liquefaction Flow Failure Case Histories, Canadian Geotechnical Journal, 39 (3): 629-647.

Rocscience, Inc. (2020), Slide2, Toronto, ON. M5T 1V1, CA.

Stark, T. D., MacRobert, C. J., and Hayter, S. A. (2020), Liquefied Strength Ratio for Eight

Laboratory-Tested Sandy Soils, Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 147 (2): 04020165.

USGS (2017), USGS Earthquake Hazards Program website, U.S. Seismic Design Maps, http://earthquake.usgs.gov/designmaps/us/application.php, accessed October 2, 2020, U.S. Geological Survey.

Whipple, J. W. (1992), Geologic Map of Glacier National Park, Montana, Department of the Interior, U.S. Geological Survey.

Youd, T. L., and Idriss, I. M. (2001), Liquefaction Resistance of Soils: Summary Report from the 1996 NCEER and 1998 NCEER/NSF Workshops on Evaluation of Liquefaction Resistance of Soils, Journal of Geotechnical and Geoenvironmental Engineering, ASCE 127 (4).

http://earthquake.usgs.gov/designmaps/us/application.php

Appendix A

• Axial Foundation Capacity Charts

• Foundation Global Stability Analysis Results

• MSE Wall and Embankment Global Stability Analysis Results

Western Federal Lands Highway Division 610 East Fifth Street Vancouver, WA 98661

Project No. West - Axial Capacity FigureMT PRA GLAC 10(45) Scale

Upper McDonald Creek Bridge Replacement A1

N/A File Name:

Date:

August 2021

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000

El ev at io n (ft

Factored Axial Resistance (kips)

30" x 0.5" Pipe Pile

Service I Limit State

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000

El ev at io n (ft

Factored Axial Resistance (kips)

30" x 0.5" Pipe Pile

Strength I Limit State

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000

El ev at io n (ft

Factored Axial Resistance (kips)

30-inch Diameter Pipe Pile

Permanent Load + Cumulative Side Resistance

Total Resistance (100% Toe) - Cumulative Side Resistance Total Resistance (50% Toe) - Cumulative Side Resistance Total Resistance (0% Toe) - Cumulative Side Resistance

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000 El ev at io n

(ft

Factored Axial Resistance (kips)

30" x 0.5" Pipe Pile

Extreme I Limit State

Vancouver, WA 98661

Project No. East - Axial Capacity FigureMT PRA GLAC 10(45) Scale

Upper McDonald Creek Bridge Replacement A2

N/A File Name:

Date:

August 2021

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000

El ev at io n (ft

Factored Axial Resistance (kips)

30" x 0.5" Pipe Pile

Service I Limit…

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000

El ev at io n (ft

Factored Axial Resistance (kips)

30" x 0.5" Pipe Pile

Strength I Limit State

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000

El ev at io n (ft

Factored Axial Resistance (kips)

30-inch Diameter Pipe Pile

Permanent Load + Cumulative Side Resistance

Total Resistance (100% Toe) - Cumulative Side Resistance Total Resistance (50% Toe) - Cumulative Side Resistance Total Resistance (0% Toe) - Cumulative Side Resistance

3,055

3,075

3,095

3,115

3,135

3,155

3,175

0 500 1,000 1,500 2,000 El ev at io n

(ft

Factored Axial Resistance (kips)

30" x 0.5" Pipe Pile

Extreme I Limit State

1.501.50

W W

250.00 lbs/ft2

250.00 lbs/ft2

1.501.50Phi (deg)Cohesion (psf)Strength Type Unit Weight (lbs/ ft3)ColorMaterial Name

360Mohr-Coulomb120MD Sand, Fill

750Undrained115Firm Silt

3521Mohr-Coulomb120 MD Sand, Alluvium

450Mohr-Coulomb140Riprap

Infinite strength150Concrete

Min FSMethod Name 1.50Bishop simplified 1.50Spencer 1.50GLE / Morgenstern-Price

East

West

-80 -60 -40 -20 0 20 40 60 80 100 120 140 160 180 200 220 240 260 280 300 320

West-Static Company FHWA WFLHDDrawn By SAH File Name McDonald Creek Bridge Replacement.slmdDate…