CO FLAP US36(1) Final Geotech Report.pdf

Geotechnical Report Estes Park Downtown Loop Reconfiguration

CO FLAP 34(1) & 36(1)

Estes Park, Colorado

May 20, 2022

Submitted To:

AECOM

7595 E. Technology Way Denver, CO 80237

By:

Shannon & Wilson, Inc.

1321 Bannock Street, Suite 200 Denver, CO 80204

23-1-01372-203

ALASKA

CALIFORNIA

COLORADO

FLORIDA

IDAHO

MISSOURI

OREGON

UTAH

WASHINGTON

WISCONSIN

1321 BANNOCK STREET, SUITE 200

DENVER, COLORADO 80204

PHONE 303•825•3800

www.shannonwilson.com 23-1-01372-203

May 20, 2022

AECOM

7595 E. Technology Way Denver, CO 80237

Attn: Mr. Stephen McQuilkin, PE

RE: GEOTECHNICAL REPORT, ESTES PARK DOWNTOWN LOOP

RECONFIGURATION, CO FLAP 34(1) & 36(1), ESTES PARK, COLORADO

We are pleased to submit our geotechnical report for the above-referenced project. This letter serves as indication that our geotechnical report was prepared and completed in accordance with Shannon & Wilson’s corporate Quality Assurance/Quality Control (QA/QC) Procedural Manual, dated May 2015. The enclosed report summarizes subsurface conditions encountered in a subsurface exploration program, laboratory tests, and geotechnical engineering design and construction recommendations.

We appreciate the opportunity to be of service to you on this project. If you have any questions or require further information, please contact me at 303-825-3800.

Sincerely, SHANNON & WILSON, INC.

David Asunskis, P.E.

Associate

DAA:/

01372-203-L1

01372-203-R1.docx 23-1-01372-203 i

TABLE OF CONTENTS

Page

1.0 INTRODUCTION

2.0 SCOPE OF WORK

3.0 SITE AND PROJECT DESCRIPTION

4.0 ROADWAY PAVEMENT CONDITIONS

5.0 SUBSURFACE EXPLORATIONS AND LABORATORY TESTING

6.0 SUBSURFACE CONDITIONS

6.1 Regional Geology

6.2 Subsurface Conditions

7.0 GEOLOGIC HAZARDS

7.1 Seismic Hazards and Ground Motion Design Parameters

7.2 Corrosive Soil and Bedrock

8.0 GEOTECHNICAL RECOMMENDATIONS

8.1 Bridge Foundations

8.1.1 Drilled Shafts – Axial Resistance

8.1.2 Lateral Resistance

8.1.3 Spread Footings for Bridge Foundations

8.2 Retaining Walls

8.2.1 Lateral Earth Pressures

8.2.2 Sliding and Bearing Resistance for Retaining Walls

8.2.3 MSE Reinforcement Length

8.2.4 Embedment Depth

8.2.5 Global Stability

8.2.5.1 MSE Wall Global Stability

8.2.5.2 CIP Wall Global Stability

8.2.6 Wall Drainage

8.3 Pavement Design

8.3.1 Traffic Loading

8.3.2 Subgrade Conditions

8.3.3 Pavement Rehabilitation Evaluation

8.3.4 Subexcavation

8.3.5 Frost Damage

8.3.6 Recommended Pavement Section

9.0 CONSTRUCTION CONSIDERATIONS

TABLE OF CONTENTS (cont.)

ii

9.1 Drilled Shaft Installation

9.1.1 Drilled Shaft Installation Methods and Equipment

9.1.2 Drilled Shaft Inspection and Observation

9.1.3 Concrete Placement

9.1.4 Non-Destructive Integrity Tests

9.2 Site Preparation

9.3 Temporary Excavations and Support

9.4 Earthwork and Grading

9.4.1 Excavation

9.4.2 Subgrade Preparation and Proof Rolling

9.4.3 Fill Placement and Compaction

9.4.4 Bulking and Swelling Factors

10.0 PLAN REVIEW AND CONSTRUCTION OBSERVATION

11.0 LIMITATIONS

12.0 REFERENCES

TABLES

1 Preliminary Pavement Assessment Summary 2 Seismic Parameters for Design 3 Corrosivity Ratings Based on Soil Resistivity 4 Sulfate Exposure 5 Ivy St. and Pedestrian Bridge Recommended Deep Foundation Design

Parameters 6 Ivy St. and Pedestrian Bridge Recommended Spread Footing Design Parameters for Bearing and Sliding Resistance on Bedrock 7 Recommended Retaining Wall Design Parameters for Lateral Earth Pressures .. 14 8 Recommended Retaining Wall Design Parameters for Bearing and Sliding

Resistance 9 Truck Distribution by Vehicle Classification 10 Recommended Pavement Sections

TABLE OF CONTENTS (cont.)

iii

FIGURES

1 Vicinity Map 2 Site and Exploration Plan 3 Generalized Subsurface Profile A-A’, Proposed Ivy Street Bridge (2 Sheets) 4 Generalized Subsurface Profile B-B’, Proposed Pedestrian Bridge 5 Recommended P-Multipliers for Group Effects 6 Recommended Surcharge Loading for Temporary and Permanent Walls

APPENDICES

A Subsurface Explorations B Laboratory Test Results C Analytical Results D Boring Location Photographs E Pavement Condition Assessment Photographs F Pavement Analysis G Global Stability Analyses H Existing Geotechnical Data I Important Information About Your Geotechnical Report

01372-203-R1.docx

GEOTECHNICAL REPORT

ESTES PARK DOWNTOWN LOOP RECONFIGURATION

CO FLAP 34(1) & 36(1)

ESTES PARK, COLORADO

1.0 INTRODUCTION

This geotechnical report presents the results of our subsurface exploration program and provides geotechnical engineering recommendations for the proposed Estes Park downtown loop reconfiguration (Project) in Estes Park, Colorado. The report summarizes our subsurface explorations, laboratory testing, and geotechnical engineering studies, and presents conclusions and recommendations for design and construction of the proposed retaining walls, bridges, and pavement.

2.0 SCOPE OF WORK

Our services were conducted in general accordance with Prime Task Order No. 16 with AECOM, PO# 99693 under Master Services Agreement No. 12S-15391-CO05. We completed the following tasks for the project:

Coordinated subcontractors for drilling borings Observed, logged, and collected soil and rock samples from ten auger borings, two of which also had continuous rock core Completed laboratory testing on selected soil and rock samples from the borings Evaluated geotechnical data and completed geotechnical engineering analyses to develop conclusions and recommendations for design project structures and pavements

Evaluated subsurface conditions relative to construction of the project Prepared this report

Our scope of services did not include performing a traffic study or evaluating the presence or absence of hazardous or toxic materials in the soil, surface water, groundwater, or air, on or below or around this site, except for limited analytical testing of soil at borings SW-08a and SW- 08b where an odor of petroleum hydrocarbons was detected by field personnel. Shannon & Wilson did not attempt to evaluate to the nature or extent of petroleum-impacted soils on the site as this was not within this scope of work.

3.0 SITE AND PROJECT DESCRIPTION

The project site is located in the downtown commercial district Estes Park, Colorado about two miles west of the Big Thompson Canyon (Figure 1). The topography slopes gently from west to east toward the Big Thompson River. The project consists of the reconstruction and realignment of E. Elkhorn Avenue (Ave.), Moraine Ave., Ivy Street (St.) and E. Riverside Drive (Dr.) into one-way streets forming a loop (Refer to Figure 2). Moraine Ave. is currently U.S. 36 and E.

Elkhorn Ave. is currently U.S. 34. In the proposed loop reconfiguration, Moraine Ave. and E.

Elkhorn Ave. will comprise westbound (WB) U.S. 36 and Ivy St. and E. Riverside Drive will comprise eastbound (EB) U.S. 36. The proposed loop will consist of two traffic lanes the entirety of the loop. The Estes Park Loop will primarily be paved with asphalt concrete pavement (ACP). Also, Rockwell St. will receive a 2-inch mill and overlay. The evaluation of the proposed improvements to E. Elkhorn Ave., Moraine Ave. (north of Fall River), and Rockwell St. are not discussed in this report.

Project elements consist of a single lane roundabout installed at the Moraine Ave., Crags Dr., and Ivy St. intersection with a designated WB lane. We understand the roundabout, as well as portions of the lead-in pavement, will be composed of Portland cement concrete (PCC) pavement. To accommodate construction of the roundabout, the current grade of Moraine Ave.

will be cut 1 to 3 feet, and up to 7 feet of fill will be placed south of the existing roadway embankment west of the proposed roundabout. Also, up to 6 feet of fill will be placed in the northeast quadrant of the roundabout. We understand three retaining walls will be placed to accommodate the roundabout. Based on preliminary plans provided by AECOM, retaining wall (RW)-01 is approximately 340 feet long and located southwest of the roundabout and will retain a maximum fill height of 11 feet; RW-02 is approximately 47 feet long and is located southeast of the roundabout and will retain a maximum fill height of 9 feet, and RW-03 is approximately 25 feet long and located north of the roundabout retaining a maximum fill height of 12.5 feet.

The project also includes the reconstruction of the Ivy St. Bridge over the Big Thompson River.

The proposed bridge will be located north of the existing bridge and consist of a two-span structure 97-feet-long skewed approximately 45 degrees from the from the river channel towards the north. A single span 67-feet-long pedestrian bridge will be installed at the existing Ivy St.

Bridge location. Along the western bank of the Big Thompson River, north of the proposed bridges, an approximate 133-foot-long retaining wall with a maximum fill height of 11 feet will be constructed (designated RW-04). We understand the southern portion of the public parking lot between Ivy St. and Rockwell St. will be reconstructed with the realignment of Ivy St.

4.0 ROADWAY PAVEMENT CONDITIONS

Based on preliminary plans, overlay pavement is considered for a portion of the Morane Ave., for approximately 750 feet south of Rockwell St., and E. Riverside Dr. for 450 feet north of Rockwell St. To analyze the condition of the existing roadway pavement for rehabilitation options, we performed a visual pavement survey to evaluate the overall pavement condition and identify areas of distress and typical crack patterns. For our pavement assessment, we developed a visual rating system for a ‘windshield’ survey from a moving vehicle with frequent stops to quantify levels of distress:

Very Good: Stable, no cracking, no patching, and no deformation. Excellent riding quality. No improvements to the pavement surface are required.

Good: Stable, minor cracking, generally hairline and hard to detect. Minor patching and possibly some minor deformation evident. Very good riding quality.

Fair: Generally stable, minor areas of structural weakness evident. Cracking is easier to detect. Patched but not excessively. Deformation more pronounced and easily noticed.

Ride quality is good to acceptable.

Poor: Areas of instability, marked evidence of structural deficiency, large areas of fatigue (alligator) crack patterns, heavy and numerous patches, and very noticeable deformation. Riding quality ranges from acceptable to poor.

Very Poor: Pavement in extremely deteriorated condition. Numerous areas of instability.

Majority of section showing structural deficiency.

Crack pattern severity levels described below are based on the Federal Highway Administration (FHWA) Distress Identification Manual (2003) using a three-level distress system (low, moderate, and high severity levels). Our assessment survey was conducted on February 9, 2018.

The pavement survey summary is presented in Table 1 with reference photographs provided in Appendix E. In our opinion, the site is feasible for overlay, but extensive rehabilitation of the existing pavement will be required. Refer to Section 8.3.3 for our pavement rehabilitation recommendations.

TABLE 1

PRELIMINARY PAVEMENT ASSESSMENT SUMMARY

Roadway Observed Pavement Condition Description

Typical Distress

Reference Photographs

E. Riverside Dr.

(U.S. 36 EB)

Overall, the condition of the pavement is fair.

Low to moderate severity longitudinal and transverse cracking observed south of E. Elkhorn Ave.

Little to no distress observed between the Big Thompson River and

Rockwell St.

Figures E-1 through E-4

Moraine Ave.

(U.S. 36 WB)

Overall, the condition of the pavement is very poor.

High severity fatigue cracking is observed at the Rockwell St.

intersection and south of Davis Street.

Moderate to high severity transverse and longitudinal cracking observed from Rockwell St. to the proposed roundabout reconstruction.

Recent utility trench was installed in the WB lane.

Figures E-5 through E-9

5.0 SUBSURFACE EXPLORATIONS AND LABORATORY TESTING

A field exploration and geotechnical laboratory testing program was implemented to explore and evaluate subsurface conditions at the proposed bridges, walls, and pavement reconstruction areas. The subsurface exploration program consisted of drilling and sampling eleven borings, designated SW-01 through SW-10, with two borings completed at boring location SW-08 (designated as borings SW-08a and SW-08b). Drilling was terminated in boring SW-08a approximately at a depth of 4.5 feet when an odor of petroleum hydrocarbons was observed beneath the base course. Boring SW-08b was later drilled offset from boring SW-08a. Borings SW-06 and SW-07 are located at the proposed pedestrian bridge and Ivy St. Bridge. Boring SW- 06 was cored approximately 30 feet into rock and boring SW-07 was cored approximately 16 feet into rock. The approximate locations of the borings are shown on Figure 2. Appendix A presents a discussion of the drilling and sampling procedures used to complete the borings.

Appendix A also presents the individual exploration logs and an explanation of the symbols and terminology used on the logs.

Geotechnical laboratory tests were completed on selected samples retrieved from the borings to estimate soil and rock index and engineering properties. Tests included natural water content, grain size distribution, Atterberg limits, unconfined compressive strength, corrosion, and R- Value. Laboratory test methods and results are provided in Appendix B. The natural water content, fines content, and Atterberg limits are also shown on the individual boring logs included in Appendix A. Appendix C contains the analytical environmental test results for soil samples from boring SW-08a and SW-08b.

6.0 SUBSURFACE CONDITIONS

6.1 Regional Geology

Broadly, the project area is located within the Front Range uplift, a mountainous area characterized by exposures of Precambrian metamorphic and intrusive igneous rocks on the order of 1.4 to 1.7 billion years old. Below the extent of glaciations, the Estes Park valley bottom is underlain by thin unconsolidated alluvial (river-deposited) and glacial outwash deposits.

Note that geologic exposures are generally covered by pavement or other structures due to the urbanized conditions along the alignment. The following description of project geology is based on review of geologic maps by Braddock and Cole (1990). This map indicates that bedrock in the approximate northern two-thirds of the project area consists of Precambrian age (approximately 1.7-billion-year-old) biotite schist, a metamorphic rock characterized by strongly developed foliation (mineral banding). In the approximate south one-third of the project area, bedrock consists of intrusive igneous rocks of the Precambrian (approximately 1.4-billion-year-old) Silver Plume Granite batholith.

Braddock and Cole (1990) indicate a surficial deposit of alluvium underlying the north part of the project area in the valley of the Fall River and the Valley of the St. Vrain River below its confluence with the Fall River. Braddock and Cole (1990) describe the alluvium as consisting of unconsolidated gravel, sand, and silt.

6.2 Subsurface Conditions

The borings completed for this project were completed through existing pavement. The existing pavement sections observed in Moraine Ave., Ivy St., and E. Riverside Dr. varied from 3.5 to 7.5 inches of ACP over 2 to 8 inches of crushed aggregate base (CAB) (where encountered). Refer to Table A-1 in Appendix A for a summary of the pavement thicknesses in our subsurface explorations. CDOT completed a pavement subsurface investigation along E. Elkhorn Ave. in 2017. The investigation limits started at Moraine Ave. (Mile Post [MP]1.306) and continued east beyond the E. Riverside Drive intersection (MP 1.43). The existing pavement thicknesses and subgrade conditions encountered are provided in Appendix H. The data suggest that the WB

ACP thickness was 4 to 7 inches thick and the EB ACP thickness was 7 inches, overlying CAB that varied in thickness for each direction.

Pavement subgrade soils (approximately 1 to 2 feet below the pavement section) generally were similar throughout the Project and consisted of a medium dense to very dense, silty sand to poorly graded sand with silt alluvium (or fill generated from alluvium). These subgrade conditions are consistent with subsurface conditions encountered below E. Elkhorn Ave.

Borings SW-07 and SW-08a encountered loose to very loose sand. American Association of State Highway and Transportation Officials (AASHTO) classification indicated typically A-2-4 to A-1-b soil types. Information was not available for the original construction of any of the existing roads.

Borings at proposed retaining walls (borings SW-01 through SW-02, and SW-09) encountered similar alluvial soils overlaying bedrock. Boring SW-01 encountered medium dense, silty sand (A-2-4) over granite bedrock at approximately 5 feet. Boring SW-09 encountered medium dense to dense, poorly graded sand with silt and gravel (A-1-a) over highly weathered gneiss bedrock at approximately 6 feet. Boring SW-02 encountered medium dense, silty sand (A-2-4) over poorly graded gravel with silt and sand (A-1-a) at 5 feet, followed by completely weathered granite at approximately 12 feet.

For the two proposed bridges, generalized subsurface profiles A-A’ and B-B’ are located respectively at the proposed Ivy St. Bridge replacement and proposed pedestrian bridge sites (refer to Figures 3 and 4). Borings near approximate bridge abutments (SW-06 through SW- 08b) also encountered alluvial soils. Boring SW-06 (near southwest abutment of the proposed Ivy St. Bridge and west abutment of the proposed pedestrian bridge) encountered a medium dense to very dense, well-graded gravel fill (A-1a) over medium dense, poorly graded gravel with silt and sand (A-1-a). Bedrock was encountered at approximately 7 feet and consisted of low to high strength, foliated to schistose gneiss. The gneiss consisted of very close to medium spaced high angle joints mostly parallel to foliation and was highly weathered from 7 to 11 feet, moderately weathered from 11 to 18 feet, and slightly weathered from 18 to 39 feet (the bottom of boring SW-06). Boring SW-07 (near the eastern pedestrian bridge abutment) encountered very loose to medium dense, silty sand with gravel (A-1-b) over dense, poorly graded sand with silt and gravel (A-1-a). Gneiss was encountered at approximately 13 feet and was similar to that of SW-06 although, highly weathered gneiss was not encountered in this boring.

The bedrock at the bridge sites was moderately weathered from 13 to 21 feet and slightly weathered from 21 to 30.7 feet (the bottom of boring SW-07). Generally, we observed that the degree of weathering was consistent with rock quality designation (RQD) and joint spacing in both borings SW-06 and SW-07.

Boring SW-08b (located near the northeast abutment of the proposed Ivy St Bridge) encountered a loose, poorly graded sand with silt and gravel (A-1-a). At 9.5 feet, very dense, poorly graded gravel with silt and sand was observed to a depth of 13 feet. Gneiss bedrock was logged from cuttings and inferred from drill action at a depth of 13 feet. The boring was terminated at a depth of 15 feet.

7.0 GEOLOGIC HAZARDS

7.1 Seismic Hazards and Ground Motion Design Parameters

Colorado is comprised of areas of low to moderate potential for damaging earthquakes. It is not possible to accurately estimate the timing or location of future earthquakes, because the occurrence of major earthquakes is relatively infrequent and the historical earthquake record in Colorado is short (about 140 years). The closest known active fault near the project area is the Williams Fork Mountains Fault. It is a northwest striking, northeast dipping, high angle normal fault located approximately 50 miles southwest of the project area (Witkind, 1976). It forms the northeastern most extension of the Rio Grande Rift. Based on geomorphic features along the fault trace, this fault is suspected to have been active less than 15,000 years ago (Kirkham, 2004).

Liquefaction may occur in loose, saturated, and cohesionless soils when subjected to earthquake ground shaking. Based on the loose to very dense sand and gravel encountered at the project site, and the relatively low peak ground acceleration (PGA) for this area, the liquefaction potential is low.

The AASHTO site class is determined using SPT N-values measured in the upper 100 feet of the soil profile. Using the average N-values within the upper 100 feet, Site Class D is recommended for the proposed bridges and retaining walls based on criteria in the AASHTO Load and Resistance Factor Design (LRFD) Bridge Design Specifications (2017).

Ground motion parameters were determined for the project site using the USGS U.S. Seismic Design Maps Web Application (USGS, 2018) and the 2009 AASHTO LRFD Bridge Design Criteria (the design parameters have not changed in the current 2017 manual). A summary of the seismic design ground motion parameters are provided in Table 1.

TABLE 2

SEISMIC PARAMETERS FOR DESIGN

Ground Motion Parameters Value Peak Ground Acceleration1 (PGAB) 0.067 g Site Class D Short-period Spectral Acceleration, Ss 0.141 g Long-period Spectral Acceleration, S1 0.036 g Site Factor, Fpga 1.6 Site Factor, Fa 1.6 Site Factor, Fv 2.4 Peak Design Spectral Acceleration, AS 0.108 g Short-period Design Spectral Acceleration, SDS 0.226 g Long-period Design Spectral Acceleration, SD1 0.086 g T0 0.076 s TS 0.381 s Seismic Zone A

Notes: 1) PGAB= peak ground acceleration for a site underlain by Site Class B rock.

2) Seismic parameters developed in accordance with AASHTO Guide Specifications for LRFD Seismic Bridge Design (2017).

7.2 Corrosive Soil and Bedrock

Soil and bedrock materials can be corrosive to substructure elements. To assist in estimating the corrosion potential at the site, samples of the overburden were tested for pH, resistivity, water soluble sulfates, and chlorides. The results are presented in Table B-1 in Appendix B.

The resistivity measured in the three samples was 6,100, 5,200 and 14,000 ohm-cm. Based on the 2018 Federal Highway Administration (FHWA) Manual “Drilled Shafts: Construction Procedures and LRFD Design Methods”, the resistivity values suggest a nonaggressive subsurface environment for metal in contact with these materials. Roberge (1999) also provides corrosivity ratings based on soil resistivity (Table 3). The resistivity values measured in the samples indicate mildly to moderately corrosive subsurface conditions.

TABLE 3

CORROSIVITY RATINGS BASED ON SOIL RESISTIVITY

Soil Resistivity (ohm-cm) Corrosivity Rating

> 20,000 Essentially noncorrosive 10,000 – 20,000 Mildly corrosive 5,000 – 10,000 Moderately corrosive 3,000 – 5,000 Corrosive 1,000 – 3,000 Highly corrosive

< 1,000 Extremely corrosive

The concentration of water soluble sulfates measured in the four samples from the site during this study were less than 0.01, 0.02, 0.03, and 0.03 percent by weight. Based on classifications as defined by ACI-318 (ACI, 2019) (Table 4), these test results suggest a negligible degree of sulfate attack on concrete exposed to site soils (exposure class S0).

TABLE 4

SULFATE EXPOSURE

Water Soluble Sulfate in Soil (Percent by Weight) Sulfate Exposure

< 0.10 Negligible (S0)

0.10 – < 0.20 Moderate (S1)

0.20 – 2.00 Severe (S2)

> 2.00 Very Severe (S3)

8.0 GEOTECHNICAL RECOMMENDATIONS

8.1 Bridge Foundations

We understand that conventional drilled shafts or continuous spread footings founded on bedrock are being considered for foundation support for the proposed pedestrian and Ivy St. bridges.

Regardless of the selected foundation type(s), we anticipate difficult installation conditions due to the presence of shallow groundwater and the strength of bedrock. Design recommendations for drilled shafts and spread footings are presented in the following sections.

8.1.1 Drilled Shafts – Axial Resistance

The design criteria presented herein uses LRFD criteria that were developed based on recommendations presented in the AASHTO LRFD Bridge Design Specifications (AASHTO, 2017). Axial resistance parameters, which are summarized in Table 5, were developed based on methods described in FHWA Manual, “Drilled Shafts: Construction Procedures and Design Methods” ( (Brown and others, 2018).

The nominal unit side resistance values were estimated from Equation 10-22 in the 2018 FHWA manual. The resistance calculated by this equation utilizes a reduction factor based on RQD and the Geologic Strength Index (GSI). The nominal unit base resistance values were estimated from Equation 10-29 in the 2018 FHWA manual. This equation provides a method of calculation for shafts bearing on intact or massive rock (Brown and others, 2010).

Drilled shafts can be designed for tip and side resistance in the bedrock. Consistent with local practice, we recommend ignoring side resistance in the overburden. The side resistance should also be ignored above the potential scour depth.

We recommend a minimum diameter of 12 inches for the proposed drilled shafts. We understand that smaller diameter drilled shafts can have advantages in terms of drilling resistance in high strength gneiss bedrock depending on the means and methods of drilling. Refer to Section 9.1.1 for a discussion of feasible excavation equipment for the site. Per our discussions with AECOM, we understand that 24-inch diameter drilled shaft are under consideration for both the proposed bridge foundations.

The drilled shaft axial resistance parameters presented in Table 5 are nominal values. For service and extreme event limit state conditions, a resistance factor of 1.0 should be applied to the nominal resistance values. The recommended resistance factors for strength limit state conditions are provided in Table 5. The nominal axial resistance parameters presented herein do not require reduction due to shaft group action, provided the shafts are spaced a horizontal distance of at least two shaft diameters, center-to-center.

Assuming that drilled shafts are constructed using good installation techniques and equipment based on criteria described in the 2018 FHWA manual (Brown and others, 2018), we anticipate that total settlement for drilled shafts at the service limit state will be approximately ¼ inch. We anticipate that differential settlement between piers will be about 50 to 75 percent of this value. About 75 percent of the settlement is anticipated to occur as the load is applied with the remaining settlement occurring within the first year following construction. Drilled shaft construction considerations are discussed further in Section 9.1.

8.1.2 Lateral Resistance

Lateral loads acting on the structure from wind, seismic, and other loadings are typically resisted by the passive earth pressure against the caps, the frictional resistance developed between the sides of the cap and surrounding soils, and the lateral resistance provided by the deep foundation members. The lateral behavior of the shafts is highly dependent on the degree of fixity of the top of the shaft.

Frictional sliding resistance at the base of the cap should be ignored because a deep foundation-supported structure may not transmit load directly to the soil beneath the cap.

Similarly, passive soil resistance against the cap and frictional resistance along the sides of the cap should be ignored because of the relatively small allowable design deflections and the potential loss of support against the cap during a scour event. As a result, the lateral resistance will come from the deep foundation elements.

We assume that the commercial software LPILE by Ensoft, Inc. (2019) will be used to evaluate the lateral behavior of the drilled shafts, including estimating lateral deflections and determining required embedment depths. LRFD design methods for lateral loading of drilled shafts utilize a resistance factor of 1.0 and are controlled by allowable movement criteria.

Geotechnical parameters required for input into LPILE are presented in Table 5. These input parameters are for an individual drilled shaft. If groups of drilled shafts are used, the effects of group interaction should be considered when evaluating the horizontal load-deflection behavior.

When using the P-Y method of analysis, the values of P (load) should be factored appropriately to account for the group effects. Recommended values of P-multipliers for loading perpendicular and parallel to a line of shafts are provided on Figure 5.

TABLE 5

IVY ST. AND PEDESTRIAN BRIDGE RECOMMENDED DESIGN PARAMETERS

Location (Boring ID)

ELEVATION 2,3

Soil/Rock Description

DRILLED SHAFT AXIAL

RESISTANCE PARAMETERS LPILE PARAMETERS FOR LATERAL ANALYSIS 4,5,6,7,8,9,10

Top (ft)

Bottom (ft)

Nominal Unit Side

Resistance, fs (ksf)

Nominal Unit Base Resistance, qb (ksf) LPILE Soil Type

Effective Unit Weight, γ'

(pcf)

Drained Friction Angle, φ'

(deg)

Compressive Strength

(psi)

Initial Modulus of Rock Mass

(psi)

Rock Quality Designation

Rock Mass Strain, εrm

West Abutments

(SW-06)

7,531 7,526 Fill - - Sand (Reese) 125 32 - - - -

7,526 7,524 Fill/Alluvium - - Sand (Reese) 63 32 - - - -

7,524 7,510 Bedrock 16.0 115 Weak Rock 103 - 900 20,000 30 0.0005

7,510 7,492 (BOE) Bedrock 36.0 750 Strong Rock 103 - 24,500 - - -

East Pedestrian Bridge Abutment

(SW-07)

7,531 7,526 Fill - - Sand (Reese) 125 32 - - - -

7,526 7,518 Fill/Alluvium - - Sand (Reese) 63 32 - - - -

7,518 7,510 Bedrock 16.0 115 Weak Rock 103 - 900 20,000 30 0.0005

7,510 7,500 (BOE) Bedrock 36.0 750 Strong Rock 103 - 24,500 - - -

East Ivy Street Bridge Abutment and Pier (SW-08b)11

7,529 7,526 Fill - - Sand (Reese) 125 32 - - - -

7,526 7,515 Fill/Alluvium - - Sand (Reese) 63 32 - - - -

7,515 7,510 Bedrock 16.0 115 Weak Rock 103 - 900 20,000 30 0.0005

7,510 7,500 Bedrock 36.0 750 Strong Rock 103 - 24,500 - - -

Notes: ft = foot; psi = pounds per square inch; ksf = kips per square foot; pcf = pounds per cubic foot; deg = degrees; BOE = bottom of exploration

1. Boring elevations were estimated from existing contour plans provided by AECOM.

2. Groundwater elevation of 7,526 feet was assumed based on the river elevation during drilling in based on observed groundwater elevation in borings SW-06 and SW-07. The design should consider flood stage of the Big Thompson River.

3. Factored shaft end bearing resistance should be calculated by multiplying the nominal unit end bearing (qb) by the end area of the shaft and a resistance factor of 0.5. Factored shaft side resistance should be calculated by multiplying the nominal unit side resistance (fs) by the side surface area of the shaft in contact with the rock within each layer and by a resistance factor of 0.55. Total factored axial compressive capacity for the shaft is determined by summing the factored end bearing resistance and the factored side resistance. Factored uplift resistance should be calculated by multiplying the total nominal side resistance over the embedded shaft length by a resistance factor of 0.45 (Brown and others, 2018). If a non-redundant single shaft is used, the above resistance factors should be reduced by 20 percent (AASHTO, 2014). A resistance factor of 1.0 is appropriate for the Service and Extreme Event Limit States.

4. The nominal shaft resistance parameters above are based on a single shaft and do not consider group action of closely spaced shafts (closer than 2 diameters, center to center).

5. Side resistance has been neglected in overburden soil above bedrock. Additionally, as stated in the report, side resistance should be ignored to whatever depth temporary casing is installed.

6. A minimum bedrock penetration (rock socket) of 3 feet is assumed for each layer. Side resistance should be ignored in the top 3 feet of bedrock due to the potential for disturbance from casing for axial resistance of drilled shafts

7. The LPile parameters shown are for a single shaft. For shaft group effects, see Figure 5 for recommended P-multipliers.

8. The LPile parameters shown are for a horizontal ground surface. Sloping ground surface modifications should be included, as necessary, per Ensoft Inc.'s recommendations for the LPile program (2019).

9. Lateral analyses should consider potential loss of ground due to scour.

10. A resistance factor of 1.0 should be used for lateral analyses.

11. Due to drilling limitations of boring SW-08b, bedrock was not cored. The boundary between weak and strong rock parameters was selected at elevation 7,510 feet based on borings SW-06 and SW-07. The boring was terminated at a depth of 15 feet.

01372-203-R1.docx

8.1.3 Spread Footings for Bridge Foundations

We understand that spread footings are also a potential alternative for foundation design. The nominal bearing resistance was calculated using Hoek-Brown criteria for the bedrock (AASHTO, 2017). We assumed no overburden for bearing resistance (scour to bedrock)

TABLE 6

IVY ST. AND PEDESTRIAN BRIDGE RECOMMENDED DESIGN PARAMETERS FOR

SPREAD FOOTING BEARING AND SLIDING RESISTANCE ON BEDROCK

Foundation Design Parameters Recommended

Value 1,2,5

Buoyant Unit Weight, γ’ (pcf) 103

Nominal Gross Bearing Resistance t (ksf):3,4,6 30 Nominal Coefficient of Friction for Sliding (tan δ): 6 0.7

Notes:

1. The parameters are applicable for continuous footings founded on directly on bedrock.

2. See AASHTO (2017) Section 11.10.7.1 and 11.6.5 for various load combinations and static forces to be evaluated.

3. Footing weight should be included when calculating bearing load.

4. A minimum footing width of 2 feet is required for bearing resistance.

5. Lateral loads assumed to be resisted by sliding.

6. Based on AASHTO (2017), we recommend the following strength limit state.

Sliding Resistance: 0.80 Bearing Resistance: 0.45

8.2 Retaining Walls

Four retaining walls, identified as RW-01 through RW-04 on the 70% plans (AECOM, 2018), indicate a maximum wall height of about 12.5 feet. We understand the walls may consist of either mechanically stabilized earth (MSE) or cast-in-place concrete (CIPC) cantilevered walls.

Currently, walls RW-01 and RW-03 are proposed MSE walls and RW-02 and RW-04 are proposed CIPC walls. Recommendations for both wall types are provided below.

8.2.1 Lateral Earth Pressures

The lateral earth pressures against retaining walls depend on many factors, including surcharge loads, the type of adjacent soil, drainage provisions, and whether or not the top of the wall can yield or deflect laterally/rotate during and after excavation. Recommended lateral earth pressure parameters for the walls were developed based on recommendations by AASHTO (2017). Our recommended earth pressure parameters are provided in Table 7. Additional recommendations and assumptions are as follows:

We recommend Structural Backfill conforming to the requirements provided in Section 704.04 of the Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects, FP-14 (FP-14) (FHWA, 2014).

Active and at-rest earth pressures provided in Table 7 assume the walls are backfilled with properly compacted (see Section 9.4.3), Structural Backfill in a 1H:1V (horizontal to vertical) zone extending upward from the base of the wall from a point

1.5 feet behind the wall.

Where walls are free to deflect, they can be designed for the active condition. Where a deflection greater than 0.001 times the height of the wall (0.001H) cannot occur, they should be designed for at-rest conditions.

The earth pressures in Table 7 assume the walls have a vertical wall face and horizontal back slope, and subsurface drainage is provided such that hydrostatic pressures do not develop.

If the conditions and assumptions above are not met, we should be notified so that we may revise our recommendations.

TABLE 7

RECOMMENDED RETAINING WALL DESIGN PARAMETERS

FOR LATERAL EARTH PRESSURES

Backfill Design Parameters

Structure Backfill or Equivalent Native Granular Backfill2

Unit Weight, γ (pcf) 135

Effective Friction Angle, φ′ (degrees) 34

Effective Cohesion, c' (psf) 0

At-Rest Earth Pressure Coefficient, K0: 0.44

At-Rest Equivalent Fluid Density (pcf): 1 (RW-01 through RW-03) 59 Buoyant At-Rest Equivalent Fluid Density (pcf): 3 (RW-04) 94 Active Earth Pressure Coefficient, Ka: 0.28 Active Equivalent Fluid Density (pcf): 1 (RW-01 through RW-03) 38 Buoyant Active Equivalent Fluid Density (pcf): 3 (RW-04) 83

Notes:

1. Equivalent non-buoyant fluid density values assume drainage provided such that hydrostatic pressures do not develop in the retained zone.

2. A horizontal backslope was assumed.

3. Design elevation for buoyant lateral pressure should be provided by hydraulic engineer or others.

Surcharge loads from motor vehicles/traffic and construction equipment will induce additional lateral loads on retaining walls. The pressures provided in Table 7 do not account for surcharge loads. Lateral loads due to various types of surcharges may be calculated by multiplying the surcharge by the at-rest earth pressure coefficient (K0) or active earth pressure coefficient (KA), depending on if the wall is free to yield as discussed above. The earth pressure coefficients are also shown in Table 7. Surcharge loads should be applied as recommended on Figure 6.

Seismic earth pressures should be added to the static earth pressures acting on retaining walls and structures. Seismic earth pressures should be calculated in accordance with AASHTO (2017). The seismic parameters for calculation of seismic stability are provided in Table 2.

8.2.2 Sliding and Bearing Resistance for Retaining Walls

The coefficient of friction for sliding resistance is provided in Table 8. The recommended nominal bearing resistance also can be obtained from Table 8 for effective retaining wall footing widths between 3 and 5 feet. The bearing resistance has been provided for the strength limit state and a service limit states for ½ and 1 inch of settlement for retaining wall foundations bearing on soil. For foundations bearing on bedrock, the anticipated movement is less than ½ inch. We recommend the service limit state bearing resistance be used for design.

Resistance factors are also provided in Table 8.

TABLE 8

RECOMMENDED DESIGN PARAMETERS FOR BEARING AND SLIDING RESISTANCE

Foundation Design Parameters Recommended Value 1,2,5 Unit Weight of Soil, γ (pcf) 115 Nominal Gross Bearing Resistance for Soil (ksf):3,4,6 8 Service Bearing Pressure for 0.5 in. of total settlement for Soil (ksf): 2.5 Service Bearing Pressure for 1 in. of total settlement for Soil (ksf): 5 Nominal Coefficient of Friction for Sliding for Soil (tan δ): 6 0.4 Buoyant Unit Weight of Bedrock, γ’ (pcf) 103 Nominal Gross Bearing Resistance for Bedrock (ksf):3,4,6 30 Nominal Coefficient of Friction for Sliding for Bedrock (tan δ): 6 0.7

Notes and assumptions:

1. The parameters are applicable for the proposed MSE walls and CIPC walls.

2. See AASHTO (2017) Section 11.10.7.1 and 11.6.5 for various load combinations and static forces to be evaluated.

3. Footing weight (including weight of soil above footing) should be included when calculating bearing load.

4. A minimum footing length of 3 feet was assumed for bearing resistance.

5. Lateral loads assumed to be resisted by sliding.

6. Based on AASHTO (2017), we recommend the following strength limit state factors for design of the retaining walls.

Sliding Resistance: 0.80 (cast-in-place), 1.0 (MSE wall) Bearing Resistance: 0.45 (cast-in-place) and 0.65 (MSE wall)

The bearing resistance values provided assume horizontal ground in front of the wall footing. At the time this report was prepared, we understand retaining wall types may change based on constructability and bridge foundation types.

8.2.3 MSE Reinforcement Length

To reduce potential for compound stability to control the design of MSE walls, we recommend a minimum L/H ratio (reinforcement length/wall height) of 0.7, or a minimum reinforcement length of 8 feet, whichever is greater. The reinforcement lengths may need to be increased to meet internal, external (sliding and overturning), or compound stability requirements. These failure modes should be evaluated by the MSE wall vendor/designer as these failure modes depend on the particular reinforcement type and spacing selected by the wall vendor/designer. The reinforcement lengths may also need to be increased to meet global stability. As discussed below, global stability analyses should be completed once the details of the walls are finalized.

8.2.4 Embedment Depth

Where the ground surface in front of the wall face is near horizontal, the base of the reinforced zone (MSE walls) or the base of the wall (CIPC walls) should be a minimum of 36 inches below the ground surface. Although not anticipated, if the ground surface in front of the wall is sloped, the bottom of the reinforced zone (MSE walls) or bottom of the wall (CIPC walls) should be located a minimum of 36 inches below the elevation at which there is a 4-foot horizontal distance from the wall face to the slope face in front of the wall. A 4-foot-wide horizontal bench should also be provided in front of walls bearing on sloping ground. The walls should be embedded a minimum of 36 inches below the ground surface for frost mitigation.

8.2.5 Global Stability

Based on the U.S. Department of Transportation and FHWA Federal Lands Highway Project Development and Design Manual (PDDM), dated March 2008, we recommend a minimum Factor of Safety (FS) of 1.5 for critical walls and 1.3 for other walls (U.S. DOT and FHWA, 2014). We recommend a FS of 1.5 for RW-04 within 50 feet of Ivy St. Bridge. Per AASHTO Section 11.5.4.2, a seismic analysis is not required if the adjusted peak ground acceleration is less than 0.4g.

8.2.5.1 MSE Wall Global Stability

Our global stability analyses assume the proposed MSE and gravity block walls should be designed and constructed in accordance with the PDDM and AASHTO criteria.

The following AASHTO MSE wall requirements and assumptions were used to develop the cross sections used in our global stability analysis:

AASHTO Section 11.10.2.2 requires a 4-foot horizontal bench in front of the MSE wall.

AASHTO Section 11.10.2.2 requires MSE wall leveling pad to extend below frost depth (36 inches). AASHTO also provides a minimum embedment depth for the reinforced zone based on proposed toe slopes and proposed height. However, based on the proposed wall layouts, the 36-inch frost embedment depth controls.

AASHTO Section 11.10.2.1 requires a minimum reinforcement length equal to 70 percent of the wall height, with a minimum length of 8 feet.

We assumed horizontal backslope and foreslope.

The intent of our stability analysis was to analyze the global stability (i.e., we did not consider failure surfaces through the reinforced zone). The MSE wall designer may need to increase the strap lengths to meet internal, external (sliding and overturning), or compound stability requirements. Based on our analyses, a reinforcement length to height ratio of 0.7 will achieve the required FS. See appendix G for our slope stability runs.

8.2.5.2 CIP Wall Global Stability

The proposed CIP walls provided in 70 percent plans did not include dimensions.

We assumed a heel length of 3 feet for our global stability analyses. Based on our analyses provided in Appendix G, a heel length of 3 feet meets the required factor of safety. We anticipate that larger heel lengths may be required, particularly for RW-04 which will include hydrostatic loads.

8.2.6 Wall Drainage

Based on conditions encountered in the subsurface explorations (see Section 6.2), we do not anticipate the presence of permanent groundwater near the base of the walls, with the exception of RW-04 and potential high flow periods in the Big Thompson River. Nevertheless, surface water can infiltrate wall backfill, regardless of the permeability of the backfill.

Therefore, we recommend providing drainage measures that reduce the potential for water to accumulate in the backfill and for hydrostatic pressures to act on the wall face. Typical drainage measures include free-draining backfill and a drainage system at the base of the reinforced and retained fill zones. The specific drainage measures should be provided by the MSE wall vendor/designer. We also recommend free-draining backfill be placed in front of the wall.

Based on plans provided by AECOM, RW-04 will be located half embedded within the Big Thompson River. We recommend analyzing the retaining wall with a hydrostatic load on the retained side and no phreatic surface and full scour on the riverside. RW-04 should also consider impact loading on the riverside due to debris during flooding or other extreme events.

8.3 Pavement Design

E. Elkhorn Ave. and Moraine Ave are State Highways (SH) 34C and SH 36A, respectively, and therefore pavement designs for proposed improvements to the Estes Park downtown loop reconfiguration are based on CDOT design procedures. However, we understand that FP-14 paving material specifications will be used during construction (FHWA, 2014). For our design, we used the procedures outlined in the 2021 CDOT M-E Pavement Design Manual (PDM) (2020) and the CDOT M-E Pavement Design Manual 2021 Addendum (Addendum) (2021) using Version 2.3.1 of the AASHTOWare Pavement M-E Design (Pavement M-E) software (AASHTO, 2013). We understand at the completion of the Project, the entire loop will be designated as SH 36. Pavement design for the parking area is designed based on procedures presented in the 1993 AASHTO Guide for the Design of Pavement Structures (AGDPS), with guidance from the PDDM (FHWA, 2008).

8.3.1 Traffic Loading

To perform a mechanistic-empirical (M-E) pavement design, detailed traffic loading information is required for the analysis. Specifically, a design annual average daily truck traffic (AADTT) and a distribution of truck vehicle type is required.

The Project traffic study prepared by Felsburg Holt & Ullevig (FHU) in 2016 provided average weekday and average weekend traffic (during peak season) volumes for three locations in the loop reconfiguration at years 2018 and 2040. In addition, the report indicated that the study was based on traffic composed of 2 to 3 percent ‘heavy vehicles’. For our analysis, we assumed a 3 percent of the of the total traffic will be composed of trucks. To determine an annual average daily traffic (AADT) volume, we assumed a weighted average of the peak seasonal weekend and peak weekday traffic volumes based on the following assumptions:

Assume the ‘peak summer’ weekend traffic will be applied for two days a week for approximately 5 months (approximately 40 days over a 365-day period).

The commuter volume will then be applied for every other day. This assumes the offseason weekend days are similar traffic volume to weekdays.

Use the higher traffic loading (E. Elkhorn Ave.) for the pavement design of the loop reconfiguration:

The source for the traffic loading data, as well as our AADTT calculations, are provided in Appendix F. Based on the Projected loop reconfiguration traffic volumes, we estimate a design one-way AADTT volume of 330 trucks per day for the initial year service (2023) with a compounded growth rate of approximately 1.41 percent. In accordance with CDOT design procedures, a 90 percent lane distribution factor was assumed for the proposed two-lane traffic loop.

For the proposed Moraine Ave., Ivy St. and E. Riverside Dr. roundabout, AECOM provided an additional traffic study using FHU’s traffic projections to estimate the peak summer weekend and average weekday traffic volumes for each quadrant of the roundabout as well as the Moraine Ave to WB US-36 bypass. Their study is provided in Appendix H. To estimate the AADTT, we performed a similar analysis as described above. The peak traffic loading occurred in the Moraine Ave to WB US-36 bypass. To design the pavement for the roundabout, bypass, and the approach PCC pavements, we analyzed two traffic loading scenarios (a) WB US-36 consisting of an AADTT of 330 trucks per day distributed over two lanes and (b) a one-lane traffic AADTT value of 231 trucks per day corresponding to the estimated bypass traffic loading.

For the roundabout, a traffic speed of 15 miles per hour was assumed.

In addition, AECOM provided us traffic counts taken over a four-day period during the peak season (Saturday, May 30, 2015 to Tuesday, June 2, 2015). The combined traffic counts for the 4-day period of E. Elkhorn Ave., Moraine Ave., and E. Riverside Drive were used to determine the truck traffic distribution (vehicle classes 4 through 13). Table 9 summarizes the vehicle distribution used in our M-E analysis for both the proposed Estes Park loop configuration and the roundabout. The 2015 traffic counts are provided in Appendix F.

TABLE 9

TRUCK DISTRIBUTON BY VEHICLE CLASSIFCATION

Vehicle Classification

Percentage Truck Traffic

Vehicle Classification

Percentage Truck Traffic

4 6.6 9 2.3 5 53.6 10 1.4 6 17.7 11 0.9 7 0.7 12 0.2 8 16.4 13 0.2

CDOT and the PDDM do not provide guidance for traffic loading for parking areas.

Based on the observed usage during the July 2013 scoping site visit (occurring during the peak season), we assume the traffic loading for the sections of parking lot that will be reconstructed will be composed primarily of automobiles with the occasional light truck. Based on guidance from the Colorado Asphalt Paving Association (CAPA), the 18-kip equivalent single axle loading (ESAL) loading for these can for this traffic ranges approximately 50,000 to 100,000.

For our analysis, we assumed a design traffic loading of 50,000 which is also the minimum ESAL value in the PDDM.

8.3.2 Subgrade Conditions

The subgrade strength for the project was evaluated by a Hveem stabilometer (R-value) test on two of the poorest quality subgrade observed in our pavement subsurface investigation (borings SW-01 and SW-04). Test results were 55 and 28 percent, respectively. Based on these results, we assumed a value of 30 percent, and used a correlation presented in the CDOT PDM to estimate the subgrade resilient modulus of approximately 8,700 pounds per square inch (psi).

8.3.3 Pavement Rehabilitation Evaluation

We recommend that prior to placing the overlay, areas of distress should be identified where the pavement should be repaired. In our opinion, the repair areas should be identified prior to milling and then re-evaluated after the milling process.