Final Geotechnical Report April 2021.pdf

BUFFALO CROSSING BRIDGE

APACHE-SITGREAVES NATIONAL FOREST

GREENLEE COUNTY, AZ

AZ FS 24(1)

FINAL Geotechnical Report

Report No. AZ-FX-0024-20-01

Prepared by Federal Highway Administration

Central Federal Lands Highway Division

Geotechnical Services Branch April 2021 ii

SIGNATURE SHEET

Report prepared by: ___________________________________________________________

Devin T. Dixon, P.E., Geotechnical Engineer

Report reviewed by: ____________________________________________________________

James M. Arthurs, P.E., Ph.D., Geotechnical Engineer

Approved for distribution by: ____________________________________________________

Marilyn D. Dodson, P.E., Lead Geotechnical Engineer

Distribution

Electronic:

N:\AZ\24(1)\Geotech\7_FinalDocs K:\TechServices\Geotech\1. Geotech Project Reports Project Management Project Development, Lead Designer

Page | i

TABLE OF CONTENTS

SECTION ONE - INTRODUCTION

1.1 BACKGROUND AND LOCATION

1.2 SCOPE AND PURPOSE

SECTION TWO - GEOLOGY AND SEISMICITY

2.1 REGIONAL GEOLOGY

2.2 SITE GEOLOGY

2.3 REGIONAL SEISMICITY

2.4 SEISMIC DESIGN PARAMETERS

2.5 GEOLOGIC HAZARDS

SECTION THREE - SUBSURFACE INVESTIGATION

3.1 SUBSURFACE EXPLORATION PROGRAM

3.2 LABORATORY TESTING PROGRAM

3.3 SUMMARY OF SITE CONDITIONS

3.3.1 General Topography

3.3.2 Surface Reconnaissance

3.3.3 Subsurface Conditions

3.3.4 Groundwater

3.4 ROCK CHARACTERISTICS

3.4.1 Generalized Strength Criterion for Intact Rock

3.4.2 Rock Characterization

SECTION FOUR - ANALYSIS & RECOMMENDATIONS

4.1 FOUNDATIONS

4.1.1 Bridge Foundation Selection

4.1.2 Site Characterization

4.1.3 Drilled Shaft Axial Resistance

4.1.4 Group Effects on Axial Resistance

4.1.5 Lateral Loads on Deep Foundations

4.1.6 Settlement

4.1.7 Field Testing

4.1.8 Scour Potential and Erosion

4.2 ABUTMENT &WINGWALL DESIGN

4.3 EARTH RETAINING STRUCTURES

4.3.1 Design Considerations

4.3.2 Gabion Wall Design

4.4 PERMANENT EARTHWORKS

4.4.1 Embankment Construction

4.4.2 Shrink/Swell Recommendations

4.5 CONSTRUCTION CONSIDERATIONS

4.6 SPECIFICATIONS

4.7 DISCLAIMER/LIMITATIONS CLAUSE

SECTION FIVE - REFERENCES

Page | ii

TABLES

TABLE 2.1 Summary of Nearby Quaternary Faults TABLE 2.2 Summary of Seismic Parameters Corrected for Site Class C TABLE 3.1 Summary of the Field Exploration Program TABLE 3.2 Summaries of Laboratory Index Test Results TABLE 3.3 Summaries of Laboratory Corrosivity Results TABLE 3.1 Summary of Uniaxial Compressive Strength Results TABLE 3.5 Strength Parameters Calculated for Rock Based on Hoek Brown Criterion TABLE 4.1 Proposed Bridge Foundation Location TABLE 4.2 Shaft Tip Resistance in Rock TABLE 4.3 LPILE Parameters TABLE 4.4 Recommended Number of CSL Access Tubes by Shaft Diameter TABLE 4.5 Lateral Earth Pressures for Bridge Abutments & Wingwalls TABLE 4.6 Recommended Safety Factors for Static Design TABLE 4.7 Material Properties Used in the Analysis for Gabion Wall TABLE 4.8 Gabion Wall Geometry TABLE 4.9 Summary of Calculated Factors of Safety for Static Conditions

FIGURES

FIGURE 1 Existing Rock Basket Structure FIGURE 2 Cross Section of Gabion Retaining Wall (Conceptual)

PLATES

PLATE 1 Project Vicinity Map PLATE 2 Geology Map PLATE 3 Geology Map Legend PLATE 4 Geotechnical Boring Locations

APPENDICES

APPENDIX A Field Exploration Program APPENDIX B Laboratory Test Results APPENDIX C Photographs APPENDIX D Excavation Characteristics of Rock & Rippability Charts

Page | 1

Buffalo Crossing Bridge, AZ FS 24(1) April 2021

SECTION ONE - Introduction

1.1 BACKGROUND AND LOCATION

This report presents the results of the geotechnical engineering study and provides recommendations for the replacement of the Buffalo Crossing Bridge in Greenlee County, Arizona. The project proposes to replace the existing Buffalo Crossing Bridge, which is located within the Apache-Sitgreaves National Forest. A project vicinity map is presented on Plate 1 of this report.

The Buffalo Crossing Bridge was constructed in 1962 and designed by 1957 AASHTO Bridge Specifications. The superstructure consists of a cast-in-place concrete deck with a haunched pier section. The substructure consists of vertical concrete cantilever abutment walls on spread footings extending to approximately 17 feet deep as indicated on the as-built drawings. Concrete wingwalls are present and rest on spread footings. Solid concrete wall piers are also founded on spread footings at approximately 19.5 and 17 feet for piers 1 & 2, respectively. All foundations are noted as founded on rock and have been reported as stable according to the most recent inspection report.

This inspection report, dated December 2018, stated that the bridge did not meet the current load requirements. The outdated bridge specifications, as noted earlier, also did not include seismic design requirements.

The bridge has seen increased logging truck traffic in recent years and the capacity has deteriorated due to both the loading and age. The Arizona Department of Transportation (ADOT) announced a permanent rule in March 2019 that increased the allowable weight limits from 80,000 pounds to 90,800 pounds on some state highways. This rule applied to the National Forest System as well and caused a number of bridges to be rated poor or deficient. Therefore, replacement of the Buffalo Crossing Bridge was determined to be the desired option, as opposed to maintenance, due to the limited remaining lifespan of the existing bridge and the new load requirements.

The proposed replacement structure type is assumed to be a simple span prefabricated steel structure on deep foundations. Deep foundations are proposed due to the higher costs anticipated with providing shallow foundations at the depth to rock near 17 feet. Additional materials, such as concrete, rebar, excavation, backfill, and dewatering, as well as possible increased construction time, would be required to construct shallow foundations. Considerations related to the abutments for the Buffalo Crossing Bridge are presented in Section 4 of this report.

1.2 SCOPE AND PURPOSE

The scope of work included a geotechnical investigation, analysis, and recommendations for bridge foundations for use in design and construction. This involved several tasks including field

Page | 2 reconnaissance, subsurface sampling, laboratory testing, interpretation and correlation of field measurements, and geotechnical engineering analysis. Specifically, this investigation was conducted to determine the soil profiles at the bridge location and develop recommendations concerning bridge foundations, retaining structures, geologic hazards, and construction considerations.

We understand that a hydraulics report has been prepared under separate cover by the CFLHD Hydraulics Engineer. The hydraulics report includes discussion of flood impacts, such as bridge scour and water surface elevations.

Page | 3

SECTION TWO - Geology and Seismicity

2.1 REGIONAL GEOLOGY

The project is located within the Mexican Highland Section of the Basin and Range Province. This province is part of the Intermontane Plateaus physiographic region. The Basin and Range Province extends from eastern California to central Utah and from southern Idaho to the state of Sonora in Mexico. This province is characterized by steep climbs up mountain ranges alternated with long treks across flat basins. Features and landscapes within this province include pediments, alluvial fans, bajadas, bolsons, Inselbergs, playas, mud flats, salt flats, lakes, sand dunes, canyons, and the Rio Grande Rift (NPS, 2020).

Quaternary and Tertiary rocks and deposits that exist within the region include siltstone, sandstone, shale, limestone, conglomerate, basalt, tuff, agglomerate, rhyolite, andesite, dikes, plugs, sills, alluvium, gravel, sand, and silt. Cretaceous aged units include the Mesa Verde group, Mancos shale, Dakota sandstone, diabase, and granite and related crystalline intrusive rocks. Jurassic, Triassic, and Permian units include the Morrison formation, San Rafael group, Glen Canyon Group, Chinle formation, Shinarump conglomerate, Moenkopi formation, Kaibab limestone, Coconino sandstone, and Supai formation. Older units, Carboniferous and Devonian to older Precambrian aged, include quartzite, Mescal limestone, Apache group, schist, and granite and granite gneiss (Wilson & Moore, 1958; Wilson et al., 1960).

2.2 SITE GEOLOGY

Although not mapped, the soil conditions at the site are assumed to be alluvium based on the stream bed deposits. The project site is mapped as underlain by basalt. The basalt locally includes tuff and agglomerate. This observation is consistent with the borings as will be discussed in Section 3 of this report. The deposits and landforms are volcanic in nature and the terrain can be described as mountainous and rugged. A geology map of the project area and corresponding map legends are presented in Plates 2 & 3.

2.3 REGIONAL SEISMICITY

No seismic hazard faults are located within 50 miles of the project. However, six known Quaternary faults are mapped within approximately 50 miles of the bridge site. These faults are considered inactive and are summarized in Table 2.1. The Vernon Fault Zone, Concho Fault, Coyote Wash Fault, and Red Hills Faults are located north and northeast of the project site and the Alma Mesa Faults and Unnamed Faults East of Alma are located south and southeast of the project site. All fault systems are considered normal faults, with the exception of Vernon Fault Zone which is considered to be left lateral, with slip rates less than 0.008 inches per year. No known active faults underlie or are closely associated with the project site (USGS, 2020a).

Page | 4

Table 2.1 - Summary of Nearby Quaternary Faults

FAULT OR

FAULT ZONE

DISTANCE

FROM

CENTER OF

PROJECT

SLIP-RATE

CATEGORY

DIP

DIRECTION

AVERAGE

STRIKE

FAULT

LENGTH

TIME OF MOST

RECENT

DEFORMATION

(miles) (inch/year) (miles) (years) Vernon Fault Zone, (Class A) No. 1016

19.1 <0.00787 NE N46°W 35.4 <750,000

Alma Mesa Faults, (Class A) No. 941

26.0 <0.00787 E; SE; W;

NW N23°E 9.32 <1,600,000

Coyote Wash Fault, (Class A) No. 1015

29.5 <0.00787 SW N42°W 26.1 <750,000

Concho Fault, (Class A) No. 1014

30.3 <0.00787 NE N37°W 24.2 <750,000

Red Hill Faults, (Class A) No. 2138

34.5 <0.00787 SE; NW N25°E 9.3 <1,600,000

Unnamed Faults East of Alma, (Class A) No. 2011

35.6 <0.00787 W; E N12°W 7.5 <1,600,000

2.4 SEISMIC DESIGN PARAMETERS

Recommended seismic response parameters for the Buffalo Crossing Bridge project site design are based on the American Association of State Highway and Transportation Officials (AASHTO) LRFD Bridge Design Specifications, 8th Edition, 2017, and represents horizontal peak ground acceleration (PGA) with 7 percent probability of exceedance in 75 years (approximate 100-year return period). The 1000-year return period uniform hazard spectrum for the Buffalo Crossing Bridge project site, located at 33.76109°N latitude and 109.35701°W longitude, was obtained in accordance with the AASHTO ground motion maps.

Based on subsurface conditions encountered during drilling and the USGS Vs30 online map viewer, the onsite soils can be assumed to be reasonably dense. Therefore, the site is classified as Class C according to the site class definitions specified in Table 3.10.3.1-1 of AASHTO. The recommended spectral acceleration coefficient values for probabilistic design are summarized in Table 2.2.

Page | 5

Table 2.2 - Summary of Seismic Parameters Corrected for Site Class C Horizontal Peak Ground Acceleration, (PGA) 0.071 g

Horizontal Response Spectral Acceleration at Period of 0.2 sec, (Ss) 0.165 g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (S1) 0.049 g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.2 Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.2 Site Factor at Long-Period Range of Acceleration Spectrum, (Fv) 1.7

Factored Horizontal Peak Ground Acceleration, (As) 0.085 g Factored Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDs) 0.198 g Factored Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.083 g

Seismic Zone Zone 1

Based on the long acceleration coefficient SD1 value of 0.083, the project site is assigned to seismic hazard “Zone 1” in accordance with Table 3.10.6-1 of AASHTO. Based on this assignment, seismic loading is not likely to control design of structures for the project.

2.5 GEOLOGIC HAZARDS

Potential geologic hazards at the bridge site include those related to floods, earthquakes, and mass movement. Flooding, however, is the most common, widespread, and damaging of these geologic hazards and could pose a threat to the Buffalo Crossing Bridge. Much of Arizona is prone to flooding caused by thunderstorms that happen year-round. Severe thunderstorms are most common during the monsoon season from mid-June through September. Flooding can cause many effects due to contact with flood waters such as extensive erosion, debris flows, and slope instability.

Earthquakes, as discussed in the previous section, are another potential geologic hazard. Eastern Arizona and the project site are located in an area of low potential for damaging earthquakes, although it is impossible to accurately predict the timing or location of future earthquakes.

Mass movements are also common in Arizona due to the steep mountainous and hilly terrain, heavy rains, and the fines content of the soil. These rock and soil movements include debris flows, landslides, and rockfalls. Although the bridge is located in terrain susceptible to mass movement, no indication of these events were observed during the site reconnaissance and are considered unlikely to impact this bridge site.

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SECTION THREE - Subsurface Investigation

3.1 SUBSURFACE EXPLORATION PROGRAM

A subsurface investigation targeting the bridge site was performed by a Central Federal Lands Highway Division (CFLHD) of the Federal Highway Administration (FHWA) geotechnical engineer on September 21 & 22, 2020. The geotechnical subsurface exploration program consisted of drilling a total of two borings each to a depth of 26-feet deep. One boring was drilled near each bridge abutment. Hollow-stem augers were used to drill through the overburden soils and bedrock samples were recovered using NQ diamond core drilling. Standard penetration testing (SPT) and sample collection was performed at 5-foot intervals for each boring when possible. Grab samples of the upper subsurface were also collected. Subsurface conditions were logged and representative samples were collected and transported to the CFLHD Materials Laboratory in Lakewood, CO, for physical property testing. Logs of the explorations and boring locations are presented in Appendix A and Plate 4, respectively. Photographs related to this exploration can be found in Appendix C. A summary of the field exploration is provided in Table 3.1.

Table 3.1 - Summary of Field Exploration Program

EXPLORATION

DESIGNATION LOCATION

APPROXIMATE

GROUND

ELEVATION (ft.)

TERMINATION

DEPTH (ft.)

DEPTH TO

GROUNDWATER

(ft.)

BH20-01

North Abutment;

STA 104+90,

8.5 ft. RT of centerline

7,541 26 Not encountered.

BH20-02

South Abutment;

STA 102+35.5,

15.5 ft. LT of centerline

7,538 26 Not encountered.

Note: The exploration locations were estimated relative to existing features. Ground elevations were estimated from Google Earth.

3.2 LABORATORY TESTING PROGRAM

Grab samples and soil samples recovered from the borings by SPT were tested in the laboratory to support the field classifications and to provide an estimate of the engineering characteristics and mechanical properties of the soil. Laboratory tests included moisture content (AASHTO T255), sieve analysis (AASHTO T 11 and T27), classification (AASHTO M145), Atterberg limits (AASHTO T89 and T90), resistivity (AASHTO T 288), and soil pH (AASHTO T 289). When the necessary tests were completed, samples were classified using the Unified Soil Classification System (USCS) and AASHTO soil classification system. Results of the testing are summarized below in Table 3.2 and are presented in Appendix B.

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Table 3.2 - Summaries of Laboratory Index Test Results

BORING

NUMBER

SAMPLE

DEPTH

PERCENT

GRAVEL

PERCENT

SAND

PERCENT

PASSING

#200

LIQUID

LIMIT

PLASTIC

LIMIT

USCS

CLASS.

AASHTO

CLASS.

(feet)

BH20-01 0 - 2 31 41 28 38 24 SC A-2-6(1)

BH20-01 4 - 5 8 64 28 31 18 SC A-2-6(0)

BH20-02 0 - 2 46 38 16 NV NP GM A-1-b(0)

BH20-02 11 - 12.5 36 4 60 28 25 ML A-4(0)

Soil test results indicated a range of material types including gravel, clayey sands, and silt, classifying as A-1-b to A-4 by AASHTO and GM, SC, and ML by USCS.

Corrosivity and chemistry testing of select samples were sent to an outside laboratory for testing and are summarized in Table 3.3.

Table 3.3 - Summaries of Laboratory Corrosivity Results

BORING

NUMBER

SAMPLE

DEPTH

RESISTIVITY PH SULFATE ION

CONTENT

CHLORIDE ION

CONTENT

(feet) (ohm-cm) (ppm) (ppm)

BH20-01 4 - 5 2,790 8.4 20 4

- Tests not completed.

Unconfined compressive strength (UCS) of rock materials was conducted on intact samples of rock core recovered from the borings. Results of the UCS testing are presented in Table 3.4.

Table 3.1 - Summary of Uniaxial Compressive Strength Results

BORING NUMBER SAMPLE DEPTH RQD1 LENGTH / DIAMETER2 UCS

(feet) (%) (psi)

BH20-01 11 20 2.4 2,030

BH20-01 21 92 2.4 1,850

BH20-01 26 92 2.5 6603

BH20-02 21 82 2.5 1,250

BH20-02 22 82 2.5 1,490

1Rock Quality Designation for core run from which sample was taken.

2Values less than 2.0 do not meet the requirements of ASTM D 7012. Not included in analysis if encountered.

3Specimens broke in less than the 2 and 15-minute requirement per ASTM D 7012. Not included in analysis.

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3.3 SUMMARY OF SITE CONDITIONS

This section presents the results of the surface reconnaissance and subsurface exploration.

3.3.1 General Topography

The general topography of the project site is characterized by steep mountains and lengthy deserts that encompass southeastern Arizona and southwestern New Mexico. Rock outcroppings are present in the area surrounding the bridge site. Thick vegetation typical of both ponderosa pine and pinyon-juniper woodland forests is abundant at the project site. The East Fork Black River flows under the Buffalo Crossing Bridge from the north to the south and converges with the West Fork Black River approximately 0.75 miles downstream of the project site.

3.3.2 Surface Reconnaissance

Existing rock basket structures are present upstream of the bridge on the west river bank. These structures supplement and mostly tie into the northeast wingwall of the north bridge abutment.

They appear to be only one basket high and are anchored by steel railroad ties installed at the front facing. Details regarding the installation of these structures is unknown at this time. A layer of rock embankment overlays the rock baskets and extends to roughly the top of the river bank. The rock baskets and rock embankment continue upstream along the length of the river in a northeast direction for at least a few hundred feet, possibly more.

Rock outcroppings protrude from a slope roughly 60 feet west of the north bridge abutment. The slope is roughly 30 feet high and appears to be around or slightly steeper than a 1V:1.5H slope ratio. Vegetation including junipers, small pines, and native grass species heavily cover much of the slope. Plate-shaped rock outcroppings protrude along the face of the slope and more block-shaped rock masses line the crest of the slope. The rock type is consistent with the geology discussed in Section 2. Much of the exposed rock faces are partially covered in moss, at least slightly weathered, and show signs of natural fracture patterns. A campground area is present on top of this slope to the northwest of the bridge.

Boulders roughly 4 feet in diameter and larger were observed in the river channel beneath and surrounding the bridge. These boulders could have been placed around the existing piers as scour protection and may have been locally sourced during the original bridge construction. Additional rock embankment surrounds the southeast bridge wingwall but is relatively shorter in length and less robust than the previously discussed embankment on the northwest river bank. Thick vegetation, as previously described, lines both banks of the river and alluvial soils and sediments encompass the bridge site. Forest Service Road 24A intersects with State Route 24 near the south abutment of the Buffalo Crossing Bridge and continues to the northeast parallel to the river channel. A gravel surfaced parking area roughly 160 feet long by 45 feet wide occupies the area just to the southwest of the bridge site. One covered U.S. Forest Service informational sign is

Page | 9 present at the parking area and a short footpath begins near this sign and terminates at the river.

Rock outcroppings again line the southwest river bank and trail area and boulders measured to be up to 3 feet in diameter were observed.

3.3.3 Subsurface Conditions

The subsurface conditions at the north abutment of the Buffalo Crossing Bridge were investigated by drilling boring BH20-01. The ground surface surrounding the boring location consisted of typical alluvial soils: namely silt, sand, and some gravel. The boring encountered sandy silt with gravel to a depth of approximately 4 feet. The material then transitioned to cobbles and basalt rock boulders less than 1 foot in diameter to a depth of approximately 15 feet. The upper 2 feet of the aforementioned cobble and boulder layer was intermixed with lean clay and gravel. The lower 9 feet of the boulders were intermixed with poorly graded sand with gravel. Tuffaceous bedrock was encountered beneath the cobbles and boulders and extended to the termination depth of 26 feet.

This rock was light red, slightly to moderately weathered, and classified as weak rock (R2).

The subsurface conditions at the south bridge abutment were investigated by drilling boring BH20-

02. Similarly, the ground surface consisted of silt, sand, and varying amounts of gravel. The boring encountered cobbles and boulders less than 1 foot in diameter intermixed with sandy silt and gravel to a depth of approximately 4 feet. The cobbles and boulders continued and became intermixed with lean clay with gravel from roughly 4 feet to 6 feet. The cobbles and boulders then became intermixed with poorly graded sand with gravel from roughly 6 to 16 feet. Tuffaceous bedrock was encountered beneath the cobbles and boulders and extended to the termination depth of 26 feet. This rock was light red, moderately weathered, and classified as weak rock (R2).

3.3.4 Groundwater

Groundwater was not encountered during drilling in either borings BH20-01 or BH20-02.

However, fluctuations in the groundwater level due to seasonal and climatic effects are expected and will likely follow water levels in the adjacent river.

3.4 ROCK CHARACTERISTICS

A natural rock mass is rarely a continuous, isotropic, homogeneous material. It can be highly variable and may be very difficult to characterize in a generalized manner. A rock mass is generally composed of intact or weathered blocks of rock separated by discontinuities such as joints or bedding planes. The discontinuity and weathering characteristics help establish the design recommendations for a foundation on rock. These characteristics include, but are not limited to, the rock lithology, strength properties, deformation properties, frequency of discontinuities (RQD), spacing of discontinuities, orientation of discontinuities, aperture, filling, condition, degree of weathering, and the groundwater condition. This information was observed at limited

Page | 10 borehole locations and laboratory testing of select rock samples. The following sections discuss the rock strength properties and rock characterization based on the field investigation and lab testing results.

3.4.1 Generalized Strength Criterion for Intact Rock

Intact samples of rock core were taken from the Buffalo Crossing Bridge boreholes and laboratory testing, such as the uniaxial (unconfined) compression test, was performed. Table 3.4 summarizes the results of the UCS testing. In the field, the rock could be scraped with a pocket knife or similar tool and it was estimated that the rock corresponded with the description of “weak rock,” R2, as described in the FHWA Rock Description Guidelines. The range of relative strength for R2 rock as listed in this reference is between 725-3,500 psi. The rock also appeared to have slight to moderate weathering.

Based on the aforementioned UCS testing, the rock strength for the Buffalo Crossing Bridge is conservatively assumed to be approximately 1,370 and 1,940 psi for abutments 1 & 2, respectively.

These values are equal to the average UCS test results at the two bridge abutments and are within the range of R2, “weak rock,” as discussed in the previous paragraph. For comparison, the cohesive strength of very stiff clay is generally between roughly 55 and 111 psi (Texas Geosciences, 2021).

3.4.2 Rock Characterization

The Geologic Strength Index (GSI) (Hoek and Marinos, 2000) system was used to characterize the rock. The GSI is a slight modification of the Rock Mass Rating system (RMR) (AASHTO, 2017) which is more appropriate for the bridge foundation evaluation at this site. The GSI is the algebraic sum of ratings assigned for the following rock mass properties:

• Intact rock strength

• Rock Quality Designation (RQD)

• Discontinuities spacing

• Discontinuities condition

• Presence of groundwater

For this project, the GSI was estimated based on collected rock core samples. The rock composition was identified as tuff with slight to moderate weathering and blocky structure. The estimated GSI values given these assumptions corresponds to a range of 60 to 70 (Hoek and Marinos, 2000). A GSI value of 65 was then assumed for the rock. Other pertinent rock properties include the Hoek-Brown constant, mi (estimated from Table 10.4.6.4-1 in AASHTO 2017), rock unit weight, γ, and the damage factor, D. These values were assumed to be 13 for tuff, 145 pcf, and 0.5 as an intermediate value for drilled shafts, respectively. For intact rock, the following equations were then used to determine the appropriate curve fitting coefficients:

Page | 11 mb = mi ∗ exp �GSI−100

For GSI > 25: α = 0.5 + 1 ∗ (exp �−GSI

� − exp �− 20 s = exp �

GSI − 100

9 − 3 ∗ D

A summary of the assumed rock properties and the Hoek Brown criteria calculations is presented in Table 3.5 below. These values are directly related to the bearing resistance discussed in Section 4.1.3.

Table 3.5 - Strength Parameters Calculated for Rock Based on Hoek Brown Criterion GSI 𝛔𝛔𝐜𝐜𝐜𝐜 (psi) 𝐬𝐬 𝐦𝐦𝐛𝐛 𝐦𝐦𝐜𝐜 𝛂𝛂

65 1,370 (abutment 1) 1,940 (abutment 2)

0.009 18.6228 13 0.502

Page | 12

SECTION FOUR - Analysis & Recommendations

This section presents analysis and recommendations for the bridge foundations, abutment wingwalls, earth retaining structures, permanent earthworks, and construction considerations for the design and construction of the Buffalo Crossing Bridge project. Based on discussions with the project team, a replacement of the existing bridge is the preferred alternative. Generalized subsurface profiles were developed based on field reconnaissance, surficial visual evaluation, and subsurface investigations.

4.1 FOUNDATIONS

The existing three span bridge will be replaced with a simple single-span structure. The existing piers will be removed and the existing abutments will remain in place. Based on the 95 percent design plans, the proposed abutment centerline station and cap elevation are shown in Table 4.1.

Table 4.1 - Proposed Bridge Foundation Location

FOUNDATION APPROXIMATE STATION TOP OF SHAFT ELEVATION*

Abutment 1 (South) 102+66 7,529.48 FT Abutment 2 (North) 103+66 7,530.44 FT

*As presented on the 95% design plans.

4.1.1 Bridge Foundation Selection

A drilled shaft foundation system is proposed for the Buffalo Crossing Bridge due to the relatively deep depth to bedrock (greater than 10 feet). Spread footers were considered but determined impractical due to additional materials, time, and cost that would be required for construction.

Spread footers would overall likely be less economical due to the size of the associated excavation and may not be able to accommodate the structure loads without being excessively large. Driven piles were also considered but determined impractical due to the presence of cobbles and boulders that could cause pile damage during driving as well as the relatively short distance, in terms of pile construction, to bedrock potentially not meeting minimum length requirements.

A drilled shaft foundation is feasible from a geotechnical perspective. Drilled shafts are able to provide a small footprint, support large foundation loads, and provide lateral resistance. The risk of scour potential is discussed in detail in the final hydraulics report. Design recommendations for the drilled shafts are provided in this report.

The following bridge loads factored for the strength limit state were provided by the CFL bridge engineer on January 8, 2020:

Page | 13

North & south abutments = 320 kips (per shaft; 3 shafts per abutment)

4.1.2 Site Characterization

The subsurface profile was assumed to be sand underlain by tuffaceous bedrock (weak, R2, bedrock) for analysis purposes. Although gravel, silt, cobbles, and boulders were encountered in the overburden materials of the designated boreholes, the soil matrix, in general, was of a sand-like nature and composition. Groundwater effects were not considered as groundwater was not encountered during drilling in either borehole.

4.1.3 Drilled Shaft Axial Resistance

Drilled shafts for the Buffalo Crossing Bridge will be socketed into tuffaceous bedrock. The side resistance provided by overburden soils and the rock socket are neglected for the shaft capacity analysis when shafts are socketed into competent bedrock. Side resistance is neglected due to potential strain-softening behavior of the sidewall, degradation of material in the borehole wall, and uncertainty of roughness in the socket sidewall. The competent bedrock is considered non-scourable; therefore, depth of scour was only considered in effects on vertical stress.

The UCS values of the competent bedrock were somewhat variable, and weaker than typical concrete mixes (Table 3.4). Due to these results and based on experience in similar projects and geotechnical materials, average values of 1,370 and 1,940 psi for abutments 1 and 2, respectively, were used for UCS to evaluate the axial capacities of the drilled shafts. These values are less than the typical strength of structural concrete and are used in accordance with AASHTO design specifications.

The shaft tip axial bearing resistance was calculated in accordance with the provisions of Article 10.8.3.5.4c. Based on the assumption of pyroclastic or otherwise blocky and fractured bedrock, the tip resistance was calculated using Equations 10.8.3.5.4c-2 and 10.8.3.5.4c-3.

𝑞𝑞𝑝𝑝 = 𝐴𝐴 + 𝑞𝑞𝑢𝑢 �𝑚𝑚𝑏𝑏

𝐴𝐴

𝑞𝑞𝑢𝑢

+ 𝑠𝑠� 𝑎𝑎

𝐴𝐴 = 𝜎𝜎′𝑣𝑣𝑏𝑏 + 𝑞𝑞𝑢𝑢 �𝑚𝑚𝑏𝑏

𝜎𝜎′𝑣𝑣𝑏𝑏 𝑞𝑞𝑢𝑢

+ 𝑠𝑠� 𝑎𝑎

In these equations, mb, s, and a are Hoek-Brown strength parameters for fractured rock masses and qu is the unconfined compressive strength of intact rock. σ'vb is the vertical stress at the tip elevation. A minimum estimated tip elevation for each foundation element has been used to calculate the nominal and factored tip resistance for various socket geometries at the strength limit state for each foundation element. This minimum tip elevation assumes a rock socket length equal to at least 150 percent of the diameter of the shaft (45 inches for a 30 inch diameter shaft). The factored resistance for the strength limit state was calculated by applying a 0.50 resistance factor

Page | 14 for tip resistance. Based on the aforementioned calculations and the selected drilled shaft diameter of 30 inches, a minimum rocket socket length of 3.75 feet (45 inches) is required to meet axial resistance. The rock socket length shown in the plans has been rounded up to 4 feet. Assumed tip elevations and resistances are summarized on Table 4.2. The surface elevations referenced on this table are located at approximately the bottom of the new abutment caps, or the top of drilled shafts, as discussed with the Bridge Engineer and noted in Table 4.1.

Table 4.2 - Shaft Tip Resistance in Rock

STRUCTURE

ELEMENT

TOP OF SHAFT

ELEV.

(feet)

ESTIMATED

SHAFT

LENGTH

ESTIMATED

TIP ELEV.

NOMINAL

TIP

RESISTANCE

(kip)

FACTORED

TIP

RESISTANCE

(kip) Abutment 1

(south) 7,529.48 18.48 7,511.00 1,583.0 791.6

Abutment 2 (north) 7,530.44 19.44 7,511.00 2,012.0 1,006.0

4.1.4 Group Effects on Axial Resistance

The resistance of a shaft group to the applied axial loads is not necessarily the sum of the axial resistance of individual shafts within the group. The zone of influence from an individual pile in a pile group may intersect with other piles, depending on the pile spacing. Historically, the efficiency of groups of drilled shafts has not been a concern as long as the center-to-center spacing between shafts is greater than three times the shaft diameter (3D) to avoid interference between adjacent shafts, assuming a single row shaft group configuration. An efficiency factor (η) should be applied for spacing less than 3D as shown in Table 10.8.3.6.3-1.

Besides the effect of overlapping zones of influence, effects of construction on ground conditions in and around the group can be significant. Excavated deep foundation elements in cohesionless soils tend to decrease the effective stress of the surrounding soils. Poorly controlled shaft construction methods can result in soil loosening during drilling and adversely reduce the lateral stress around other shafts within the group. Casing driven in advance of excavation may increase the relative density and effective stress of the surrounding soil and prevents caving of overburden material.

4.1.5 Lateral Loads on Deep Foundations

Lateral load analysis was performed by the bridge engineer using the software program LPILE developed by Ensoft, Inc. This program analyzes a single pile or shaft considering deflection as a function of design loads, foundation construction, and subsurface conditions. Table 4.3 provides preliminary LPILE input parameters for the foundation soils based on available subsurface

Page | 15 information and presumptive engineering correlations. It is also recommended that lateral support within five feet of the final road grade be neglected due to the potential loss of support from frost penetration or other shallow ground disturbance.

Table 4.3 - LPILE Parameters

APPROXIMATE

ELEVATION

(FT)

DEPTH BELOW

TOP OF SHAFT

(FT)

LPILE P-Y

MODEL

EFFECTIVE

UNIT WEIGHT

(PCF)

FRICTION

ANGLE

(DEG)

SOIL

MODULUS

(PCI)

MODEL No. 1 (SAND & STRONG ROCK)

7,529.48 to 7,515.00 (Abutment 1)

7,530.44 to 7,515.00 (Abutment 2)

0 to 14.48 (Abutment 1)

0 to 15.44 (Abutment 2)

Sand (Reese) 120 32 90

TRANSITION TO ROCK PROPERTIES

LPILE P-Y

MODEL

EFFECTIVE

UNIT WEIGHT

(PCF)

UNIAXIAL

COMPRESSIVE

STRENGTH (PSI)

7,515.00 to 7,511.00 (Abutments 1 & 2)

14.48 to 18.48 (Abutment 1)

15.44 to 19.44 (Abutment 2)

Strong Rock (Vuggy

Limestone) 145 1,655

Material properties provided are for single shafts and do not account for the reduced lateral resistance of shafts in a group. P-multipliers are a function of the number of rows of shafts and center-to-center shaft spacing in the direction of loading. P-multipliers are required even for a single row of shafts if the center-to-center spacing is less than 5 shaft diameters. P-multipliers are specified in Table 10.7.2.4-1 in AASHTO. When this analysis method is used, the resistances at the strength limit state as represented by the P-y curves should not be factored since they already represent the nominal conditions.

4.1.6 Settlement

A resistance factor of 1.0 for the service limit state is recommended to assess the ability of the foundation to meet the specified deflection criteria. Elastic settlements will occur immediately and be essentially complete at the end of construction and are estimated to be less than one inch at all locations based on the loads provided. Differential settlements are not anticipated. Shaft group settlement is not anticipated.

4.1.7 Field Testing

Steel crosshole sonic logging tubes (1.5-inch diameter steel tubes) should be installed in all drilled shafts prior to concrete placement for integrity testing. Crosshole sonic logging tubes cannot be installed in shafts less than 2 feet in diameter; therefore, a minimum shaft diameter of 2.5 feet is

Page | 16 recommended. The recommended number of access tubes and tube spacing are dependent on the selected shaft diameter; refer to Table 4.4. The crosshole sonic logging tests should be conducted in accordance with ASTM D-6760 and FP-14 for quality assurance/quality control of the drilled shafts.

Table 4.4 - Recommended Number of CSL Access Tubes by Shaft Diameter

SHAFT DIAMETER

D (ft)

NUMBER OF TUBES TUBE SPACING

(degrees)

2.5 ≤ D < 3.5 3 120

3.5 ≤ D < 5 4 90 5 ≤ D < 7 6 60

4.1.8 Scour Potential and Erosion

Contraction scour and local abutment scour depths were estimated by the hydraulics engineer.

Section 4.4.3 in the Final Hydraulic Report details the scour depth estimates.

Riprap for embankment protection is shown in the plans near the abutment wingwalls per guidance from the bridge and hydraulics engineers. For additional scour information please refer to the previously mentioned Final Hydraulic Report dated January 2021.

4.2 ABUTMENT & WINGWALL DESIGN

Abutments and wingwalls should be designed to resist lateral earth pressures and other applicable lateral loads in accordance with AASHTO. Lateral earth pressure is influenced by the strength of the abutment backfill, the presence or absence of water, and the ability of the abutment or wall to move in response to lateral loads. Other loads, such as live loads, construction loads, and soil compaction loads should also be considered in the design.

Unbalanced water behind an abutment or wall adds significant lateral pressure and should be avoided by using free draining gravity outlets for water. Abutment and wingwall backfill should consist of structural backfill as specified in Section 704.04 of FP-14.

The coefficient of at-rest earth pressure should be used for design if the abutment is so restrained that it cannot be expected to rotate (deflect at the top) 0.002 times the wall height. Where deflection of the abutment can be expected, a coefficient of active earth pressure should be used for wall design. Active and at-rest lateral earth pressures of native materials and properly placed and compacted structural backfill above the water table are presented in Table 4.5. The values are unfactored loads and assume that the surface of the soil slope behind the wall is horizontal.

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Table 4.5 - Lateral Earth Pressures for Bridge Abutments & Wingwalls

BACKFILL TYPE

ASSUMED

BACKFILL

PROPERTIES

CASE

UNFACTORED

EQUIVALENT

FLUID DENSITY

(PCF)

NOMINAL

FRICTION

FACTOR

Structural Backfill c = 0 psf φ = 34 deg.

γ = 125 pcf

Active 35 0.28

At-Rest 55 0.44

Native Soil c = 0 psf φ = 32 deg γ = 120 pcf

Active 37 0.31

At-Rest 56 0.47

4.3 EARTH RETAINING STRUCTURES

The designer has proposed to raise the roadway grade leading to the north bridge abutment by approximately 7 feet. Adjacent to the northeast bridge abutment is a designated wetland area that is prohibited from environmental impact and harm, as well as an existing rock basket structure of questionable integrity. This existing structure, as mentioned in Section 3.3.2, can be seen in Figure

1. Therefore, it has been proposed to construct an earth retaining structure at the northeast quadrant of the bridge to retain the new embankment material from the bridge approach (roadway) grade raise, allow widening for the proposed guardrail sections, and to avoid impacts to the protected wetland. The following sections discuss the design alternatives that were considered as well as the design approach and methodology for the chosen earth retaining structure.

Figure 1 - Existing Rock Basket Structure

Page | 18

4.3.1 Design Considerations

Six design alternatives were considered to address the aforementioned roadway grade raise and conflict with the adjacent wetland area. These alternatives were considered to be the most economical and practical for the project and the Forest Service. These six design alternatives were as follows:

• Extend the proposed concrete bridge wingwalls further & found on either drilled shafts or spread footers below the scour depth.

• Steepen the fill slope leading to the bridge, if possible by design standards, and maintain revegetation.

• Construct a reinforced soil slope (RSS).

• Construct a gabion retaining wall.

• Construct a gabion face MSE wall.

• Place gabion baskets on top of existing rock basket structures.

The gabion retaining wall was ultimately chosen as the most cost effective option and would have the lowest potential impact to the wetland. The other options would presumably have larger footprints which would risk disrupting the environmentally sensitive areas. The existing rock basket structures will be removed and riprap will be placed at the base of the proposed gabion wall to protect the wall from scour. The following section discusses the gabion wall design.

4.3.2 Gabion Wall Design

Gabion walls are earth retaining structures constructed using large twisted wire baskets filled with rock that are stacked, typically without connections between the baskets. These structures rely on the weight, size, shape, and interface friction of the individual baskets to resist earth pressures and provide overall stability. Gabion walls are frequently used due to their lower cost and improved aesthetics compared to traditional concrete retaining walls. A cross section of a typical FHWA gabion retaining wall is shown in Figure 2.

Page | 19

Figure 2 - Cross Section of Gabion Retaining Wall (Conceptual) [provided by Eastern Federal Lands Highway Division, Detail E253-01]

The gabion wall was designed based on traditional earth pressure theory using the GAWACWin software published by Maccaferri, Inc. The software uses limit equilibrium theory to analyze sliding, overturning, eccentricity, bearing, and global stability. Limited surficial geotechnical investigations were conducted near the wall site; therefore, a conservative design approach was followed. Presumptive soil strength properties for retained and foundation soils were assumed for observed on-site soils. The gabion wall was designed as a gravity retaining wall and a surcharge load of 250 pounds per square foot was applied to the backfill to account for traffic loading. Static lateral earth pressures were applied to the back of the wall; however passive earth pressure was neglected at the toe of the wall. As stated in Section 2.4, due to the classification of the site as Seismic Hazard Zone 1, seismic loading was not considered in the design.

Page | 20

The design was checked for adequate factors of safety against sliding, overturning, and bearing capacity. The recommended factors of safety for static gabion wall design are presented in the table below:

Table 4.6 - Recommended Safety Factors for Static Design

MODE OF FAILURE STATIC FACTOR OF SAFETY

Sliding 1.5 Overturning 2.0

Bearing 3.0 Global 1.3

Assumed material properties used in the design of the gabion walls are presented in Table 4.7.

Stability analysis of the wall did not include hydrostatic pressures behind the gabions. The gabions and the wall backfill are assumed to be free draining, such that elevated water levels in the wall backfill will not occur. Wall geometries and the calculated wall design factors of safety for each wall analysis will be discussed in the following paragraphs.

Table 4.7 - Material Properties Used in the Analysis for Gabion Wall

ASSUMED

LOCATION

φ

RETAINED

SOIL

FRICTION

ANGLE

ɣS

RETAINED

SOIL UNIT

WEIGHT

c

SOIL

COHESION

µ

SLIDING

FRICTION

FACTOR

ɣR

GABION

FACE UNIT

WEIGHT

STA 103+82 to STA 103+35 32° 120 pcf 0 psf 0.625 125 pcf

The gabion wall should be constructed starting at the end of the northeast bridge wingwall near STA 103+82 with an offset of approximately 18.4 feet. The gabion wall should follow the approximate angle of the bridge wingwall and terminate around STA 103+35. At this location, the toe of the embankment slope has been eroded by intermittent stream flow. At their maximum, the exposed height of the gabion wall is estimated to be 9 feet. The base of the wall should also be embedded a minimum of approximately 2 feet. The base width will be determined by the height of the proposed wall at each location. The backslope angle, β, should not be greater than 26.6 degrees. The gabions should be constructed in accordance with Special Detail 253-B, including a 12V:1H face batter and 1-foot set back at courses where the width changes. An underdrain is not necessary due to the free-draining nature of gabions.

Page | 21

The gabion baskets exposed to the stream flow should be coated with polyvinyl chloride (PVC), epoxy, or other inert materials to reduce the potential for damage to the wall structure. Gabions exposed to the site soils, not treated with a coating should be galvanized and provided with sacrificial wire thickness to account for corrosion during the service life of the wall. Non-woven geotextile should be placed against the back of the gabions to prevent backfill soil from migrating into the baskets. The gabion wall should be backfilled with properly compacted structural backfill material. The toe of the gabion wall should be embedded below the scour depth identified by the CFLHD hydraulics engineer. Table 4.8 summarizes the required base widths with a given wall height and Table 4.9 shows the related factors of safety.

Table 4.8 - Gabion Wall Geometry

H

MAXIMUM HEIGHT

B

MINIMUM BASE WIDTH

6 ft 6 ft 9 ft 9 ft

12 ft 9 ft 15 ft 12 ft 18 ft 12 ft 21 ft 15 ft

Table 4.9 - Summary of Calculated Factors of Safety for Static Conditions

MAXIMUM

HEIGHT

SLIDING

FACTOR OF

SAFETY

OVERTURNING

FACTOR OF

SAFETY

BEARING

CAPACITY

FACTOR OF

SAFETY

STATIC GLOBAL

STABILITY

FACTOR OF

SAFETY

6 ft 2.5 3.5 3.6 1.4 9 ft 2.1 3.7 5.3 1.4

12 ft 2.1 4.2 6.8 1.5 15 ft 1.9 3.0 4.0 1.4 18 ft 2.0 3.5 5.1 1.5 21 ft 1.8 2.7 3.6 1.4

The contractor will verify the limits of the structure and submit gabion drawings according to FP- 14, Section 253 Gabions and Revet Mattresses and 104 Control of Work.

Page | 22

4.4 PERMANENT EARTHWORKS

4.4.1 Embankment Construction

Embankment construction will be necessary with regards to the proposed roadway grade raise and concrete wingwalls. Construct permanent long-term embankments with a maximum slope ratio of 1V:2H to maintain slope stability and promote slope vegetation.

4.4.2 Shrink/Swell Recommendations

On-site soils expected to be encountered within the project limits generally consist of clayey sand and silt with varying amounts of gravel. It is estimated that these soils will have a 10 percent shrink percentage, corresponding to a shrink/swell factor of 0.90. The recommended shrink/swell factor is based on a combination of standard tabled values for common materials in the FLH Technical Guidance Manual (2006) and experience with other CFLHD projects in similar materials.

4.5 CONSTRUCTION CONSIDERATIONS

Roadway Excavation: Excavate using equipment capable of removing the material while preventing material from escaping outside the construction limits.

Based on the subsurface investigation and surface geologic mapping, the rock encountered, primarily basalt and tuff, is expected to be rippable in the area surrounding the abutments of the Buffalo Crossing Bridge; however, final determination of bedrock rippability is the responsibility of the contractor. Shear wave velocity of the geologic layers was not evaluated by geophysical methods. However, based on Table D.1 in Appendix D, the rock encountered can be presumed to be between the soft and hard rock descriptions. Additional information regarding the excavation characteristics of rock and rippability charts can also be found in Appendix D.

Evidence of definitive in-place bedrock was encountered from the subsurface investigation detailed in this report. Appropriate construction equipment should be mobilized to the site based on the rock information detailed above. The contractor is responsible for safety of excavations and shoring design.

Drilled Shafts: The contractor will be required to submit a drilled shaft construction plan according to Section 565.04 of the FP-14, which includes outlining the proposed methods to maintain borehole stability, manage the excavation of rock, concrete placement, and dewatering.

Tuffaceous rock was encountered at all the exploration boreholes. Be prepared for cobbles and boulders ranging in size from approximately 12 to 48 inches and larger. Loose subsurface materials were also encountered at shallow depths, which may warrant the use of temporary casing.

Page | 23

It is anticipated that groundwater could be encountered at the foundation excavations. The likelihood of flowing sands could also present a challenge for the driller as the mixture of fine sand and water could infiltrate the drilling equipment. The contractor should be prepared for wet construction methods as groundwater can infiltrate even when temporary casing is used.

Corrosivity: Based on guidelines outlined in subsection 704.08(b) of FP-14, a pH below 5 or above 10, a resistivity below 3,000 ohm-cm, and sulfate and chloride concentrations above 200 and 100 parts per million, respectively, are representative of an aggressive soil environment. The sample tested at boring BH20-01 from 4-5 feet resulted in a resistivity of 2,790 ohm-cm. This test result is not within the guidelines defined in subsection 704.08(b) of FP-14 and minor caution should be taken with regards to corrosion potential.

Dewatering: Dewatering may be needed due to water infiltration from the river. Perform dewatering according to Section 208.07 of the FP-14.

Slope Instability: Slopes could become unstable during construction operations due to precipitation, flooding events, or heavy loading on soft subgrade soils. Precautions should be taken if movement or cracking is observed. Seepage is a contributor to slope instability and similar dewatering and stabilization methods could be used to suspend slope movement.

Hard Layers: Hard layers, such as rock, were encountered…