3.1_-_Murray_Street_Bridge_Final_Foundation_Report_July_2023.pdf

FINAL FOUNDATION REPORT

Murray Street Bridge Seismic Retrofit and Barrier Replacement

City of Santa Cruz, California Crawford File No. 2009-0272-3

Bridge No. 36C0108

Prepared by:

Crawford & Associates, Inc.

1100 Corporate Way, Suite 230 Sacramento, CA 95831

July 2023

Prepared for:

TRC

10680 White Rock Road, Suite 100

Rancho Cordova, CA 95670-6061

Corporate Office: 4701 Freeport Boulevard | Sacramento, CA 95822 | (916) 455-4225 Modesto: 1165 Scenic Drive, Suite B | Modesto, CA 95350 | (209) 312-7668 Pleasanton: 6200 Stoneridge Mall Road, Suite 330 | Pleasanton, CA 94588 | (925) 401-3515 Rocklin: 4220 Rocklin Road, Suite 1 | Rocklin, CA 95677 | (916) 455-4225 Ukiah: 100 North Pine Street | Ukiah, CA 95482| (707) 240-4400

July 17, 2023 Crawford File No. 2009-0272-3

Cameron Pinkerton, PE

TRC

10680 White Rock Road, Suite 100 Rancho Cordova, CA 95670-6061

Subject: FINAL FOUNDATION REPORT

Murray Street Bridge Seismic Retrofit and Barrier Replacement City of Santa Cruz, California

Dear Mr. Pinkerton, Crawford and Associates, Inc. (Crawford) is pleased to submit this Final Foundation Report associated with the Murray Street Bridge Seismic Retrofit and Barrier Replacement Project located in the City of Santa Cruz, California. Crawford prepared this report in accordance with our agreements between TRC and Crawford (previously Taber Consultants). This report contains the results of our subsurface exploration and conclusions and recommendations for design of proposed new bridge foundations and approach roadway sections. Foundation recommendations for the appurtenant retaining walls are reported under a separate cover. This report updates and supersedes our Draft Foundation Report, dated January 13, 2016.

Thank you for the opportunity to be part of your design team. Should questions arise concerning foundation conditions at the site, please do not hesitate to call on us. An opportunity to review and comment on plans and specifications insofar as they rely on this report is an integral part and a condition of our recommendations.

Sincerely, Crawford & Associates, Inc., Reviewed by:

Keiko Lewis, PE W. Eric Nichols, PE, CEG Senior Engineer Principal

FINAL FOUNDATION REPORT Crawford Murray Street Bridge Seismic Retrofit and Barrier Replacement File: 2009-0272-3 City of Santa Cruz, California July 17, 2023

TABLE OF CONTENTS

1 INTRODUCTION

1.1 PURPOSE

1.2 GEOTECHNICAL SERVICES

2 PROJECT DESCRIPTION

3 GEOTECHNICAL INVESTIGATION

3.1 DRILLING AND SAMPLING – TABER CONSULTANTS

3.1.1 1997 EXPLORATION

3.1.2 2013 EXPLORATION

3.1.3 2015 EXPLORATION

3.1.4 LOGGING

3.1.5 SAMPLER PENETRATION RESISTANCE (N-VALUE)

3.1.6 BOREHOLE ABANDONMENT

3.1.7 BORING LOCATIONS AND ELEVATIONS

4 LABORATORY TESTING

5 GEOTECHNICAL CONDITIONS

5.1 SITE GEOLOGY

5.2 SURFACE CONDITIONS

5.3 SUBSURFACE CONDITIONS

5.3.1 SEDIMENTARY ROCK (SANDSTONE)

5.3.2 LAGOON DEPOSITS

5.3.3 FILL/SURFICIAL NATIVE MATERIALS

6 GROUNDWATER

7 AS-BUILT FOUNDATION DATA

8 SCOUR DATA

9 CORROSION EVALUATION

10 SEISMIC DESIGN INFORMATION AND RECOMMENDATIONS

10.1 GROUND MOTION

10.1.1 SOIL CLASSIFICATION

10.1.2 RECOMMENDED SEISMIC DATA

10.2 OTHER SEISMIC HAZARDS

10.2.1 SURFACE FAULT RUPTURE

10.2.2 LIQUEFACTION POTENTIAL

10.2.3 SEISMICALLY INDUCED SETTLEMENT

10.2.4 SEISMIC SLOPE INSTABILITY

10.2.5 TSUNAMI

11 DISCUSSION AND CONCLUSIONS

11.1 GEOTECHNICAL DESIGN PARAMETERS

12 GEOTECHNICAL RECOMMENDATIONS

12.1 FOUNDATION DATA AND LOADING

12.2 FOUNDATION DESIGN RECOMMENDATIONS

12.3 PILE DATA TABLE

12.4 ABUTMENT AND BENTS – 16” AND 30” CISS PILES

12.4.1 COMPRESSIVE (AXIAL) RESISTANCE

12.4.2 TENSION (UPLIFT) RESISTANCE

City of Santa Cruz, California July 17, 2023

12.4.3 LATERAL RESISTANCE

12.4.4 PILE GROUP REDUCTION

12.4.5 NEGATIVE SKIN FRICTION

12.4.6 SETTLEMENT

12.5 ABUTMENT 10 ANCHOR PILES (96-INCH CIDH)

12.6 APPROACH FILL

12.6.1 EARTHWORK

12.6.2 FILL MATERIAL

12.6.3 EXPANSIVE MATERIAL

12.6.4 FILL PLACEMENT AND COMPACTION

12.6.5 SLOPE GEOMETRY AND STABILITY

12.6.6 EROSION CONTROL

12.6.7 SETTLEMENT

12.7 LATERAL EARTH PRESSURES (ABUTMENTS AND WINGWALLS)

12.7.1 STATIC LATERAL EARTH PRESSURE

12.7.2 SEISMIC LATERAL EARTH PRESSURE

12.7.3 SURCHARGE LOADS

12.8 STRUCTURAL SECTION AND ROADWAY SUBGRADE

13 NOTES FOR SPECIFICATIONS

13.1 GENERAL CONSIDERATIONS

13.2 DEEP FOUNDATIONS

13.2.1 DRIVEN PILES

13.2.2 DRILLED PILES

13.3 EXISTING UTILITIES

14 NOTES FOR CONSTRUCTION

14.1 PILOT HOLES

14.2 DRIVEN PILES

14.2.1 CISS PILE INSTALLATION

14.2.2 PILE DRIVING CRITERIA

14.2.3 DRILLING TO ASSIST PILE DRIVING

14.2.4 PILE REFUSAL

14.3 CIDH PILES

14.3.1 ABUTMENT 10 ANCHOR (96-INCH CIDH PILES)

14.4 VIBRATION AND NOISE MONITORING

14.5 EXCAVATION AND SHORING

14.6 DEWATERING

14.7 EXISTING FOUNDATIONS

15 RISK MANAGEMENT

16 LIMITATIONS

LIST OF TABLES

TABLE 1: RETROFIT PILES SUMMARY

TABLE 2: APPROXIMATE ROCK SURFACE ENCOUNTER

TABLE 3: SUMMARY OF GROUNDWATER DATA

TABLE 4: CORRELATED SHEAR WAVE VELOCITY

TABLE 5: POTENTIALLY LIQUEFIABLE SOIL ZONES/LAYERS

TABLE 6: SOFT AND VERY SOFT COHESIVE ZONES/LAYERS

TABLE 7: SUMMARY OF SLOPE STABILITY RESULTS

TABLE 8: ESTIMATED ROCK ELEVATIONS

TABLE 9: FOUNDATION DESIGN DATA

TABLE 10: FOUNDATION FACTORED DESIGN LOADS

TABLE 11: FOUNDATION DESIGN RECOMMENDATIONS

TABLE 12: PILE DATA TABLE

TABLE 13: SUMMARY OF THE DOWNDRAG RESULTS

TABLE 14: RECOMMENDED EQUIVALENT FLUID WEIGHT (EFW)

TABLE 15: NEW PAVEMENT STRUCTURAL SECTIONS (R-VALUE = 50)

APPENDIX I - FIGURES

Figure 1: Vicinity Map Figure 2: Locations of Field Tests Figure 3: Geologic Map Figure 4: Fault Activity Map

APPENDIX II

Log of Test Borings (LOTB) – Murray Street Bridge LOTB – Retaining Walls

APPENDIX III

Laboratory Test Results

APPENDIX IV

Ground Motion Data Sheet

APPENDIX V

Liquefaction Analysis SLIDE Output Idealized Geotechnical Parameters Foundation Design Calculations

APPENDIX VI

General Plan Foundation Plans

APPENDIX VII

Site-Specific Seismic Response Spectrum of Free-Surface Horizontal Motion for Murray Avenue Bridge

1 INTRODUCTION

1.1 PURPOSE

Crawford & Associates, Inc. (Crawford) prepared this Final Foundation Report associated with the Murray Street Bridge Seismic Retrofit and Barrier Replacement project in City of Santa Cruz, California. Services for this project were completed in accordance with the agreement, dated July 12, 2021. The purpose of this report is to provide earth materials criteria for use in design of the proposed new bridge foundations and approach roadway. Geotechnical design of foundation elements for this project is with respect to no-collapse criteria for an ordinary bridge.

This report updates and supersedes our Draft Foundation Report (Taber File No. 2009-0272-1), dated January 13, 2016. Information and comments on that study are incorporated herein, as appropriate. Geotechnical design recommendations for appurtenant retaining walls are reported under separate covers.

1.2 GEOTECHNICAL SERVICES

To prepare this report, Crawford:

• Discussed the project with the design team from TRC;

• Reviewed in-house file data regarding this project/site that included:

• Preliminary Foundation Investigation – Seismic Retrofit, dated May 5, 1999 (Taber Consultants Job Number 1R2/396/66.5);

• Preliminary Design Report – Draft, dated November 19, 2010 (Taber Consultants Job Number 2009-0272-1);

• Foundation Report – Draft, dated January 13, 2016 (Taber Consultants Job Number 2009-0272-1);

• Reviewed the following as-built plans:

• General Plan sheet for the Santa Cruz Small Craft Harbor, Project II, dated February

8, 1962, prepared by Earl and Wright Consulting Engineers;

• Sheet titled General Plan, Plan and Elevation, and Pile Plan sheets for the Santa

Cruz Small Craft Harbor, Upper Harbor, Project I Railroad Bridge, dated December 8, 1969, prepared by Earl and Wright Consulting Engineers, and;

• Sheets titled Bent 6 Modification, Plan and Elevation, Bent 2 & 3 Details, Bent 4 Details and Partial General Plan & Elevation and Foundation Plans Bents 2, 3 & 4 for the Santa Cruz Bridge Repairs [Federal Aid Project No. ER-1606(2)], dated February 22, 1990, prepared by Parsons Brinckerhoff Quad Douglas (PBQD);

• Reviewed published geologic, seismic, and tsunami maps and literature pertaining to the project site;

• Subcontracted StrongMotions, Inc. to complete a report of Site-Specific Seismic Response Spectrum of Free-Surface Horizontal Motion for Murray Avenue Bridge (Report Number: SMI-12-2010), dated August 30, 2010;

• Completed additional geotechnical test borings at the site between March 23 through 29, 2013 and February 9 through 13, 2015;

• Completed laboratory testing on soil/rock samples obtained during subsurface explorations;

• Reviewed Final Draft project plans (100% submittal) for Murray Street Bridge Seismic Retrofit and Barrier Replacement, revision date 04-21-23, prepared by TRC; and

• Performed geotechnical engineering evaluation and analysis.

Limitations of this study are discussed in the final section of this report.

2 PROJECT DESCRIPTION

The project site is located at Murray Street where it crosses Woods Lagoon in the City of Santa Cruz, California. The site coordinates at the center of the bridge are approximately latitude 36.9680° N and longitude 122.0031° W. The site location is shown on Figure 1 in Appendix I.

The project involves retrofitting the existing structure, including: widening the bridge on both sides, adding piles for structure support, widening/lengthening of supports at all abutment and bent locations, and constructing a curtain wall at Bent-9.

Proposed retrofit strategy is shown on the “Murray Street Bridge, Seismic Retrofit and Barrier Replacement” plan sheets referenced. The foundation improvements include:

• New vertical and battered 16-inch diameter cast-in-steel-shell (CISS) piles at Abutment-1;

• New battered (1:12) 30-inch diameter CISS piles at Bents-2 through 8;

• New vertical and battered 16-inch diameter CISS piles at Bent-9 and Abutment-10, and;

• New 96-inch diameter CIDH anchor piles east of Abutment-10.

Since the presence of existing foundation piles and the need to avoid constricting the navigation channel for bents in water substantially limit the utility of various foundation types, our previous study determined that more or less “optimal” retrofit performance and reasonable constructability at Bents 2 through 8 was available with the use of 30-inch diameter piling battered at 1:12 (horizontal:vertical). Vertical piling of any kind would require a much larger pile footing for retrofit. Also, very little lateral soil resistance (particularly during earthquake loading) is available and the lateral foundation stiffness will be largely derived from various battered components of piling.

The current Foundation Plan shows retrofit piles to consist of the following:

Table 1: Retrofit Piles Summary

Support Location Proposed Pile Type Number of

Vertical Piles

Number of Battered Piles /

Batter Abutment-1 CISS 16”x0.5” 4 4 / 1:5

Bent-2 CISS 30”x0.625” 0 2 / 1:12 Bent-3 CISS 30”x0.625” 1 2 / 1:12 Bent-4 CISS 30”x0.625” 0 2 / 1:12 Bent-5 CISS 30”x0.625” 0 4 / 1:12 Bent-6 CISS 30”x0.625” 0 4 / 1:12 Bent-7 CISS 30”x0.625” 0 4 / 1:12 Bent-8 CISS 30”x0.625” 0 4 / 1:12 Bent-9 CISS 16”x0.5” 4 4 / 1:12

Abutment-10 CISS 16”x0.5” 4 4 / 1:5 Abutment-10 Anchor 96” CIDH 2 0

The potential for seismically induced ground movement (see Section 10.5) at the east bank was identified and evaluated during our previous study. As determined/reported from the previous study, the maximum (ultimate) soil load transmitted to the bridge structure as a result of ground movement is estimated as 3680±kips. This load is to be largely resisted by two 96-inch diameter CIDH pile anchors located 50-feet behind Abutment-10. Previous study recommended specified pile tip at elev. -50.0 (i.e., elev. -53.0, NGVD29 datum) for the 96-inch diameter CIDH anchor piles behind Abutment-10. This was based on verification that the top of the rock surface was at/above elev. 7 ft (NGVD29 datum).

The elevations for the test borings completed in 1997 (B-1 through B-7) are based on as-built information from 1963. According to TRC, the current project datum for vertical control is the National Geodetic Vertical Datum of 1929 (NGVD 29), where NGVD 29 = 1997 – 3.0 feet (e.g., Elev. 5.6 ft 1997 = Elev. 26 ft NGVD 29). Unless otherwise noted, elevations presented in this report are with respect to NGVD 29.

3 GEOTECHNICAL INVESTIGATION

The following provides a generalized description of the drilling, sampling and logging methods and procedures used for the 1997, 2013 and 2015 field exploration and testing phases of this project completed by Taber Consultants. The boring locations are shown on Figure 2 in Appendix I. The Log of Test Borings (LOTB) drawings in Appendix II provide more specific soil/rock descriptions, elevations, boring details and explanation of engineering geology descriptors and terms used to log the rock cores.

The test borings made for this bridge circa 1962 by Earl and Wright, Inc. are shown on the As- Built Log of Test Borings included in Appendix II. The elevations shown for the borings completed by Earl and Wright, Inc. are based on the datum of mean lower low water.

3.1 DRILLING AND SAMPLING – TABER CONSULTANTS

3.1.1 1997 EXPLORATION

A total of seven auger/rotary drilled and sampled borings (B-1 through B-7) were completed by Taber Consultants from June 1 to 11, 1997. The maximum depth below ground surface explored was 113±feet (Boring B-6; lowest elev. -127±feet). Three of the 1997 borings (B-5, B-6 and B-7) were drilled from an overwater barge platform. The land-side borings (B-1 through B-4) were typically drilled with continuous solid auger until initial groundwater was encountered and then advanced by mud-rotary method to final depth. All of the borings completed on the barge were drilled by mud-rotary method. A temporary steel casing was set through the barge deck and into the channel bottom as necessary in order to achieve an adequate ground seal for drill fluid recirculation during drilling.

Soil and rock samples were recovered from all of the 1997 borings by means of a 2.0-inch outside diameter (O.D.) “Standard Penetration Test” (SPT) sampler driven into the ground with a manually operated (rope and cathead) and uncalibrated 140-pound safety hammer with 30-inch drop (350 ft-lb striking force). Drive samples were typically collected at approximate 5-foot intervals in the borings. Bulk samples of soil cuttings were also collected at selected depths from the borings.

3.1.2 2013 EXPLORATION

Between March 23 and 29, 2013, eight additional borings (A-13-008 through A-13-015) were drilled at the site to further characterize subsurface foundation conditions for the bridge and roadway approaches. One dynamic cone penetration boring (D-13-016) was also completed.

The exploratory test borings were drilled by Taber Drilling and advanced using solid auger and/or mud-rotary drilling methods through soil and rock. The underlying rock was also cored with an HQ (3.8-inch O.D.) diamond bit core barrel. The 2013 test borings were drilled with a track-mounted CME 55 drill rig.

Soil and decomposed rock samples were recovered by means of 2.0-inch O.D. SPT and/or 2.5-inch inside diameter (I.D.) split-spoon samplers advanced with standard 350 ft-lb striking force using a calibrated automatic trip hammer. The dynamic cone penetration boring (D-13-016) was similarly advanced. Thin-walled, 3-in diameter Shelby tube samples were also taken in fine-grained soil layers for laboratory testing. The hammer used for the drive samples during the 2013 subsurface exploration is recorded to have a hammer energy efficiency (ERi) of 87% based on an energy calibration completed on April 9, 2012. Drive samples were typically collected at approximate 5-foot intervals in the borings. Bulk samples of soil cuttings were also collected at selected depths from the borings.

3.1.3 2015 EXPLORATION

Between February 9 and 13, 2015, eight additional borings (A-15-017A through A-15-024) were drilled at the site to obtain subsurface data for the proposed retaining walls. The borings were drilled by Taber Drilling and advanced using solid auger and/or mud-rotary drilling methods through soil and rock. Except for Boring A-15-023, the 2015 borings were drilled with a trailer mounted CME 45 drill rig. Boring A-15-023 was drilled with portable drilling equipment and sampled with a manually operated 35-pound hammer falling 22-inches (64 ft-lb hammer energy). For reference, the Log of Test Borings for the retaining wall elements are included in Appendix II.

Soil and rock samples were recovered by means of 2.0-inch O.D. SPT and/or 2.5-inch inside diameter (I.D.) split-spoon samplers advanced with standard 350 ft-lb striking force using an automatic trip hammer. The hammer used for the drive samples during the 2015 subsurface exploration (excluding A-15-023) is recorded to have a hammer energy efficiency (ERi) of 104% based on an energy calibration completed on November 18, 2014. Drive samples were typically collected at approximate 5-foot intervals in the borings. Bulk samples of soil cuttings were also collected at selected depths from the borings.

3.1.4 LOGGING

The borings were logged and earth materials field-classified by an engineer/geologist as to consistency, color, texture, and gradation on the bases of penetration resistance, examination of samples and observation of drill cuttings. Where rock cores were obtained, the percent recovery and Rock Quality Designation (RQD) were recorded for the recovered portions of rock.

Selected portions of recovered soil/rock samples from the driven samplers were retained in sealed containers for laboratory testing and reference. The Shelby tubes were sealed at each end of the tube with bee’s wax and capped to preserve moisture content. The Shelby tubes were kept vertical as collected in the field and transported to the laboratory in an upright position. Rock cores were retained in core boxes for reference. The bulk soil samples collected from auger cuttings were placed in plastic bags for laboratory testing and reference.

Groundwater observations were also made in the borings during and at completion of drilling operations.

3.1.5 SAMPLER PENETRATION RESISTANCE (N-VALUE)

The in-situ sampler penetration resistance (N-value) in blows per foot was recorded in the field to obtain an approximate measure of the dynamic resistance of the soil. The N-value was recorded as the number of hammer blows necessary to drive the sampler the final 12-inches of the 18-inch sample interval, unless refusal was met.

The SPT N-value adjusted to 60% hammer energy (N60) is routinely used to provide an index of the apparent density of cohesionless soils and sometimes (albeit less reliably) to estimate the consistency of cohesive soils. The energy-corrected N60 value normalized for effective overburden stress referred to as (N1)60 is typically used to correlate soil strength parameters and bearing characteristics.

For a non-standard sampler (i.e., non-SPT sampler), the in-situ N-value was corrected to an Equivalent SPT N-value using values published in Caltrans Geotechnical Manual (March 2021), then adjusted to provide an Equivalent N60 and/or Equivalent (N1)60 value that can be correlated to soil strength and bearing characteristics for use in geotechnical analysis.

The in-situ (uncorrected) N-values are shown on the Log of Test Borings (LOTB) drawings provided in Appendix II.

3.1.6 BOREHOLE ABANDONMENT

The test borings for the bridge and retaining walls were backfilled with cement grout at the completion of drilling. The shorter roadway borings were backfilled with soil cuttings. The borings were also surface sealed with asphalt-concrete cold patch where drilled within paved areas.

3.1.7 BORING LOCATIONS AND ELEVATIONS

The details and locations of test borings completed by Taber Consultants are shown on the Log of Test Borings (LOTB) drawings in Appendix II. The test borings made for this bridge circa 1962 by Earl and Wright, Inc. are shown on the As-Built Log of Test Borings included in Appendix II.

Taber Consultants personnel located the borings in the field relative to existing site features and subsequently to project stationing. The boring elevations for the 1997, 2013 and 2015 field studies were surveyed in the field by Taber personnel. The datum for the elevations shown for the borings completed by Earl and Wright, Inc. are with respect to mean lower low water (datum conversion to NGVD29 unknown). Eric Hilmer, C.E.G., Glen Wade, P.G., and Amanda Kahn, G.I.T. were the field geologists for the respective 1997, 2013 and 2015 field studies.

4 LABORATORY TESTING

The following laboratory tests were performed on soil and/or rock samples obtained from the test explorations:

• Moisture Content - Dry Density (ASTM D2216 / D2937),

• Unconfined Compressive Strength (ASTM D2166),

• Particle Size Analysis of Soils (ASTM D422),

• Atterberg Limits (ASTM D4318),

• Consolidated Undrained Triaxial (ASTM D4767),

• Sulfate/Chloride Content (CTM 417/422),

• pH and Minimum Resistivity (CTM 643),

• R-value (CTM 301), and

• Point Load Index (ASTM D5731).

Laboratory test results are shown on the LOTB drawings in Appendix II and/or provided in Appendix III.

5 GEOTECHNICAL CONDITIONS

5.1 SITE GEOLOGY

In the vicinity of the Murray Street Bridge, published1 geologic mapping shows three geologic units as follows:

• Basin deposits (Holocene);

• Lowest emergent coastal terrace deposits (Pleistocene); and

• Siltstone and Sandstone – Purisima Formation (Pliocene and upper Miocene).

The higher elevations at the margins of the lagoon are mapped as “Lowest emergent coastal terrace deposits” of Pleistocene-age. Materials of this geologic unit are described as semiconsolidated, generally well-sorted sand with a few thin, relatively continuous layers of gravel. Locally it includes many small areas of fluvial and colluvial silt, sand and gravel.

Underlying this unit is siltstone and sandstone of the Pliocene and upper Miocene-age Purisima Formation. This geologic unit is described as very thick bedded yellowish-gray tuffaceous and diatomaceous siltstone containing thick interbeds of bluish-gray, semifriable, fine-grained andesitic sandstone.

The “Basin deposits” of Holocene-age are shown overlaying the Purisima Formation within the lagoon and along the lower reaches of the slopes bordering the lagoon. These deposits are described as unconsolidated, plastic, silty clay and clay rich in organic material. Locally this unit includes interbedded thin layers of silt and silty sand.

No active fault traces are shown on published mapping to cross the site. Site seismicity and information on local faults and seismic sources is presented in Section 10.2.1. A Geologic Map is shown as Figure 3 in Appendix I.

5.2 SURFACE CONDITIONS

The existing 35±feet wide by 540±feet long structure is aligned east-west and consists of nine 60-foot spans of reinforced concrete deck on precast, prestressed concrete I-girders. The substructure is comprised of two-column bents and open-style seat abutments. The structure was damaged in the 1989 Loma Prieta earthquake. Bent-2 sustained some transverse lateral displacement and as much as 2-inches of settlement. Subsequent repair work included strengthening displaced columns and excavating to visually observe pile conditions at Bents-2, - 3 and -4. Bent-6 was extended on the south side of the bridge. It is not known if damage to the railroad bridge also occurred in 1989.

Upstream (north) of the bridge, Woods Lagoon is aligned northeast-southwest; downstream (south) of the bridge, Woods Lagoon is aligned north-south. The bridge is located just

1 Brabb, Earl E., Geologic Map of Santa Cruz County, California, United States Geological Survey (USGS), Miscellaneous Investigation Series, Map I-1905, Scale 1:62,500, 1989.

downstream (south) of the confluence of a southerly flowing tributary stream that enters into Woods Lagoon from its west bank. A Santa Cruz County Regional Transportation Commission (SCCRTC) railroad bridge that also crosses over Woods Lagoon is located 20-35±feet east of the Murray Street Bridge.

Woods Lagoon is the location of the Santa Cruz Small Craft Harbor. Construction of the bridge and the harbor included placement of artificial fill along the margins of the lagoon and dredging of a boat channel below the bridge. A previous structure known as the East Cliff Drive Bridge that crossed over Woods Lagoon was also demolished/removed when the existing bridge and harbor improvements were completed.

An existing 30-inch diameter sewer force main crosses below the lagoon on the south side of the bridge. The sewer force main will be abandoned and replaced by a new sewer force main located on the north side of the bridge. Overhead utility lines are also present along Murray Street in close proximity to the existing bridge.

5.3 SUBSURFACE CONDITIONS

Three generalized units of earth materials are identified from the test borings: underlying sedimentary rock, lagoon deposits, and fill/surficial native materials. The lagoon deposits can be further subdivided into sand, bay mud and most recent channel deposits (from bank erosion and silting of dredge channel). Fill materials include bridge approach embankments and area fills placed on both sides of the lagoon as part of harbor improvements.

The following provides a generalized description of the earth materials encountered in the borings Taber Consultants completed from the bridge foundation elements of this project. The LOTB drawings in Appendix II provide more specific soil/rock descriptions, elevations, boring details and explanation of engineering geology descriptors and terms used to log the rock cores.

5.3.1 SEDIMENTARY ROCK (SANDSTONE)

As encountered in the test borings, the underlying rock unit consists of variably weathered sedimentary rock (sandstone) that is generally described as thickly to massively bedded, decomposed and intensely to moderately weathered, very soft to moderately soft and moderately to slightly fractured. From west to east along the bridge alignment, the top of the underlying sedimentary rock as encountered in the borings completed for the bridge is shown in Table 2.

Table 2: Approximate Rock Surface Encounter

Boring Number

Year Completed

Nearest Support

Top of Boring Approximate Rock Surface Encounter

Elevation (feet, NGVD29)

Depth (feet)

Elevation (feet, NGVD29)

B-4 1997 Abutment-1 29.5 16.0 +13.5 A-13-010 2013 Bent-2 7.6 40.0 -32.4

B-2 1997 Bent-2 and -3 6.8 78.0 -71.2 D-13-016 2013 Bent-3 7.0 41.0 -34.0

Boring Number

Year Completed

Nearest Support

Top of Boring Approximate Rock Surface Encounter

Elevation (feet, NGVD29)

Depth (feet)

Elevation (feet, NGVD29)

A-13-015 2013 Bent-3 and -4 7.4 14.0 -6.6 B-1 1997 Bent-4 8.0 10.0 -2.0 B-7 1997 Bent-6 2.6* 58.0 -55.4 B-6 1997 Bent-7 -0.5* 99.5 -100.0 B-5 1997 Bent-8 2.5* 100.0 -97.5

A-13-009 2013 Bent-9 7.5 71.0 -63.5 A-13-008 2013 Bent-9 8.0 51.0 -43.0

B-3 1997 Abutment-10 12.6 28.0 -15.4 Notes: The locations/depths/elevations shown in the table are approximate.

* Denotes barge deck elevations at time of drilling. Other top of boring elevations denote the ground surface at the time of drilling.

The upper portions of the underlying rock unit were penetrated with light-weight drilling equipment using tungsten-carbide tipped drill bits. Some cemented zones were highly resistant to this drilling method. Where attempted, Standard Penetration Test (SPT) samplers typically achieved only partial penetration of the rock unit. In the borings (A-13-008, A-13-009, A-13-010 and A-13-015) where the rock was cored with a diamond drill bit, the recovery varied from 54 to 100% and Rock Quality Designation varied from 41 to 93%.

Point Load Index test results on very intensely to intensely weathered rock core samples from the borings typically indicate weak rock with correlated uniaxial compressive strengths that range from 0 to 500 pounds/square inch (psi). It was noted that these samples were soft and broke apart with light pressure. One sample tested from Boring A-13-009 (Run B) indicates a correlated uniaxial compressive strength of 3,200 psi. Factors accounting for the variability in Point Load strength include rock composition, fracturing, grain size and weathering characteristics. The Point Load Index test results are included in Appendix III.

5.3.2 LAGOON DEPOSITS

The low areas in the rock surface (consistent with incised channels) are filled with natural lagoon deposits comprised of a sequence of mostly loose to medium dense (locally dense) silty sand/sandy silt with local woody and organic material and local gravel lenses overlain by typically very soft organic silty clay (“bay mud”) with shells.

At the west bank and tributary channel area (Borings B-1, -2, -4, -10, -15 and -16), lagoon deposits encountered in these borings vary in thickness and generally consist of 6±feet (A-13-

015) to 31±feet (B-2) of materials interpreted as “bay mud” underlain by sedimentary rock (sandstone) expect in Boring B-2 where the “bay mud” is underlain by about 30±feet of materials generally described as loose and very loose (locally dense) silty sand. Lagoon deposits are essentially absent in Boring B-1/Bent-4 and in Boring B-4/Abutment-1.

In the main Woods Lagoon channel (Borings B-5, -6 and -7), the uppermost 5-8±feet of soil, interpreted as recent channel deposits, is described as loose-very loose silty sand and very soft sandy silt and silt. Lagoon deposits underlying these soils vary in thickness and generally consist of 34±feet (B-7) to 47±feet (B-5) of materials interpreted as “bay mud” underlain by about 4±feet (B-7), 35±feet (B-5) and 37±feet (B-6) of materials generally described as loose to medium dense (locally very loose, compact and dense) clayey/silty sand.

At the east bank (Boring B-3, -8, -9), lagoon deposits encountered in the borings vary in thickness and generally consist of 10±feet (B-3), 45±feet (A-13-008) and 63±feet (A-13-009) of materials interpreted as “bay mud” underlain by sedimentary rock (sandstone).

Laboratory test results on fined-grained soil samples from this unit classify as silt (ML) with Liquid Limit from 21 to 29 and low plasticity (PI ≤ 5) to non-plastic (NP); lean clay (CL) with Liquid Limit from 27 to 55 and Plasticity Index ranging from 13 to 33; and fat clay (CH) with Liquid Limit from 50 to 69 and Plasticity Index ranging from 25 to 40. The water content of nearly all tested samples exceeds the Liquid Limit, consistent with salt water deposition and flocculent soil structure.

The granular soils and low plasticity silt layers of this unit are in the range of texture and consistency potentially susceptible to liquefaction; “bay mud” materials may be “sensitive”, i.e.

they may suffer substantial loss of shear strength if/as disrupted (conversely, their in-place undisturbed strength may exceed that exhibited in field and laboratory testing performed for this study). Liquefaction potential of these soils is presented in Section 10.2.2.

5.3.3 FILL/SURFICIAL NATIVE MATERIALS

At the west and east banks, materials interpreted as fill and surficial native materials as encountered in the borings completed for the bridge are generally described as loose to slightly compact clayey/silty sand and soft (locally stiff) clayey silt/silty clay with local/scattered wood, peat and other debris. Such materials were encountered from the ground surface (or below nominal asphalt pavement section) to depths of 6±feet to 18±feet. The uppermost 3±feet to 13±feet of this unit appear to be mostly fill and may be in-part naturally occurring material eroded from adjacent terraces. These materials overly the lagoon deposits except in Borings B- 1 and B-4 where they directly overly the sedimentary rock (sandstone).

6 GROUNDWATER

The groundwater conditions encountered in the borings completed for the bridge are presented in Table 3. The water surface in Woods Lagoon varies due to tidal influence.

Table 3: Summary of Groundwater Data

Boring Number

Ground Surface

Elevation (feet, NGVD29)

Depth to Groundwater

(feet)

Groundwater Elevations

(ft)

Date Measured

B-1 8.0 7.6 0.4 07/01/1997 B-2 6.8 10.8 -4.0 07/02/1997 B-3 12.6 13.2 -0.6 07/02/1997 B-4 29.5 19.0 10.5 07/03/1997

Boring Number

Ground Surface

Elevation (feet, NGVD29)

Depth to Groundwater

(feet)

Groundwater Elevations

(ft)

Date Measured

B-5 2.5 Channel Boring/Woods Lagoon 07/09/1997 B-6 -0.5 Channel Boring/Woods Lagoon 07/10/1997 B-7 2.6 Channel Boring/Woods Lagoon 07/11/1997

A-13-008 8.0 6.5 1.5 03/25/2013 A-13-009 7.5 Not Established 03/25/2013 A-13-010 7.6 6.5 1.1 03/26/2013 A-13-015 7.4 10.5 -3.1 03/29/2013

D-13-016 7.0 Not Measured/Dynamic Cone Penetration Test Boring 03/29/2013

7 AS-BUILT FOUNDATION DATA

The existing foundations are shown on construction plans by Earl & Wright for the original bridge (1962), for bridge footing additions at Bents-6, -7 and -8 (1969, with 1972 “As-Built” modifications) and bridge repair plans by Parsons Brinckerhoff Quad Douglas (PBQD, 1990).

All 1962 structure support is by means of driven piling – cast-in-place piles with 16-inch diameter shells in the upper 40-feet and 12-inch diameter shells below. Design pile load is not noted on the 1962 plans.

For 1962 bridge construction, the estimated pile lengths shown on the plans range from 30 to 40 feet at abutments to as much as 90 feet at Bent-5. A note on the 1962 plans indicate that piles at Abutment-1 through Bent-5, Bent-9, and Abutment-10 are to be “driven to rock”; other piles to be driven to specified tip elevation. Actual pile lengths installed are not noted and, based on previous test boring encounters; they are likely to vary substantially from the estimated lengths shown on the plans.

Plans for the Railroad Bridge and Murray Street Bridge bent footing modifications (1969) show 1972 “as-built” pile tip elevations. The as-built pile tip elevations for Murray Street Bridge footing modifications at Bent-6, -7 and -8 are shown on the “Pile Plan” for that project. Comparison of pile tip elevations with boring logs suggests that some of these piles achieved bearing within dense granular soil above rock. Substantial penetration of concrete piles into rock is considered unlikely.

The bottom of existing pile footings are shown for Abutment-1 at elev. +20; Bent-2 through 9 at elev. +3; and Abutment-10 at elev. +11. The bent footings are shown as constructed at then existing ground level on a fill berm across Woods Lagoon (1962); subsequently (1969), the ground was excavated to elev. -11 to -13± around Bents-5 through -9 and fill placed at/around Bent-2, -3, and -4 and between Bent-9 and Abutment-10. Some concrete encasement has also been poured around the top of piles at Bent-9.

Following the Loma Prieta earthquake, the Bent-6 pile cap was further extended to the south with the addition of four driven pre-cast concrete piles. A Bent-6 Foundation Modification Plan (dated March 1, 1990) prepared by PBQD shows the modification piles as Class 45C with 25-ton design (service) load. The pile tips are shown at elev. -60.0 (elev. -63.0, NGVD29).

However, no pile driving records (or notes) are shown on the plans and the as-built pile tip elevations are not known.

8 SCOUR DATA

The grade of the existing channel is artificially maintained by dredging; other than possible over-dredging, channel degradation is not expected. While considered remote, there is some possibility of erosion/scour (unknown magnitude) due to temporary increased flow rate through the channel due to a tsunami event (see Section 10.2.5).

9 CORROSION EVALUATION

For structural elements, Caltrans2 defines a corrosive environment as an area where the soil has either a chloride concentration of 500 ppm or greater, a sulfate concentration of 1,500 ppm or greater, or has a pH of 5.5 or less. With the exception of MSE wall design, Caltrans does not require soil and water are to be tested for chlorides and sulfates if the minimum resistivity is greater than 1,100 ohm-cm.

The project is located in a marine environment and within a tidal channel (chloride concentration in seawater is approximately 19,000 mg/l) that is considered corrosive and the design of reinforced concrete and steel foundation elements is recommended to consider potential exposure to corrosive salt water and marine atmosphere. Section 12 of the 2021 Caltrans guidelines provides information regarding corrosion mitigation measures for structural elements and lists additional Caltrans guideline documents regarding corrosion mitigation.

10 SEISMIC DESIGN INFORMATION AND RECOMMENDATIONS

10.1 GROUND MOTION

Due to the presence/thickness of soft, compressible and potentially liquefiable soil layers identified at this site, a site-specific seismic analysis was completed for this project by StrongMotions, Inc. (Sharon, Massachusetts). Their report titled “Site-Specific Seismic Response of Free-Surface Horizontal Motion for Murray Avenue Bridge, Report Number: SMI- 12-2010” dated August 20, 2010 is attached as Appendix VII.

StrongMotions Inc. used latitude 36.967998° and longitude -122.002316° for the site coordinates and an envelope correlated shear wave velocity (VS30) from firm rock to soft rock site conditions in the upper 100 feet of the soil profile equal to 760 meters per second (about 2,500 feet per second) and 474 meters per second (about 1,554 feet per second) respectively.

These value corresponds to a “rock” with 760 m/s < Vs < 1500 m/s (Soil Profile Type B) and a

2 Caltrans Corrosion Guidelines Version 3.2, May 2021

“very dense soil and soft rock” with 360 m/s < Vs < 760 m/s (Soil Profile Type C) for the upper 100 feet of the soil/rock profile and is considered appropriately conservative for use in bridge design at this site.

The VS30 value was determined for this site based on the subsurface data obtained from the test borings and correlations with SPT blow count N-values corrected for hammer efficiency using the equations outlined by Caltrans3. The correlated VS30 values estimated from the test boring logs are shown in Table 4.

Table 4: Correlated Shear Wave Velocity

Boring Designation

Support Location

Top of Boring

Elevation

(NGVD29,

feet)

Bottom of Boring

Elevation

(NGVD29,

feet)

Total Boring Depth (feet)

Correlated Shear Wave Velocity in Upper 100 feet*

VS30

(m/sec)

VS30

(ft/sec)

B-4 Abut-1 29.5 -16.8 46.3 390 1282 B-2

Bents-2 to -9

6.8 -88.5 95.3 104 342 B-1 8.0 -37.3 45.3 409 1342 B-7 2.6 -83.4 86.0 139 457 B-6 -0.5 -126.5 126.0 64 210 B-5 2.5 -113.9 116.4 59 192 B-3 Abut-10 12.6 -37.7 50.3 252 826

* The depth to “soft” bedrock (sandstone) varies from 10 feet for Boring B-1 to 87 feet for Boring B-5.

The report by StrongMotions, Inc. identified the two controlling sources for this site as the Monterey Bay-Tularcitos (Monterey Bay section) and the San Andreas Fault zone (Santa Cruz Mountains section). The Monterey Bay-Tularcitos (Monterey Bay section) has a maximum magnitude of 7.3 and is located approximately 11.8 km (7.3± miles) southwest of the site. The San Andreas Fault zone (Santa Cruz Mountains section) has a maximum magnitude of 7.9 and is located approximately 17.2 km (10.6± miles) northeast of the site.

10.1.1 SOIL CLASSIFICATION

Due to granular soil layers with (N1)60 < 30, cohesive soil layers with Su < 2,000 psf, low shear wave velocity (Vs < 886 ft/sec), and potentially liquefiable soil layers, the site is considered “Non-Competent (Class S2)” with respect to “Classification of Soils” as outlined in the April 2019 Caltrans Seismic Design Criteria (SDC) Version 2.0 with October 2019 Interim Revisions.

10.1.2 RECOMMENDED SEISMIC DATA

Based on the results of the site-specific seismic analysis completed by StrongMotions, Inc., we recommend that structure design be based on the following Caltrans SDC parameters:

3 California Department of Transportation, Methodology for Developing Design Response Spectrum for Use in Seismic Design Recommendations, Appendix A, November 2012.

• Soil Profile Type: F;

• Maximum Moment Magnitude (Mmax): 7.9;

• Peak Ground Acceleration (PGA): 0.60g; and

• Controlling Spectra: pseudo-acceleration and deformation response spectrum values of free-surface horizontal motions for the bridge for 5% of critical damping as recommended by StrongMotions, Inc. and presented in Table 9 on page 24 of their report (see Appendix VII).

The Designed Ground Motion Data Sheet is attached in Appendix IV.

10.2 OTHER SEISMIC HAZARDS

10.2.1 SURFACE FAULT RUPTURE

No active fault traces are shown on published mapping to cross the site and the site is not within or adjacent to an Alquist–Priolo Earthquake Fault Zone for fault rupture hazard. The nearest mapped fault is the Ben Lomond Fault, is located approximately 2± miles to the west of the site.

This fault is considered not active within the Holocene by the California Geologic Survey and is not included in the faults shown on the Caltrans “Seismic Hazard Map”. No evidence of surface fault rupture was observed from our geologic reconnaissance at the bridge site. On these bases, the potential for surface fault rupture within the project limits is considered low. We show nearby faults on Figure 4 in Appendix I.

10.2.2 LIQUEFACTION POTENTIAL

Liquefaction is a secondary effect associated with seismic loading that can occur when saturated, relatively loose sand and specific soft, fine-grained saturated soils are subject to ground shaking strong enough to create soil particle separation that results from increased pore pressure. This separation and subsequent pore pressure dissipation can lead to decreased soil shear strength and settlement. Liquefaction is known to occur in soils ranging from low plasticity silts to gravels. However, soils most susceptible to liquefaction are clean sands to silty sands and non-plastic silts.

Liquefaction susceptibility of a soil deposit is a function of the soil grain size, relative density, percent fines, plasticity of the fines, degree of saturation, age of deposit, and earthquake ground motion. According to Caltrans guidelines, liquefaction potential is evaluated using the “Simplified procedure” to a depth of 70 ft in the soil profile below the channel bottom. Caltrans guidelines cite Bray and Scancio (2006) criteria to determine liquefaction susceptibility of fine-grained soils when the Plasticity Index (PI) £ 12 or the [water content (wc) / liquid limit (LL)] >

0.85. Boulanger and Idriss (2006) recommend considering a soil to have clay-like behavior (i.e., not susceptible to liquefaction) when the PI ³ 7. Predominately fine-grained (cohesive) soils such as clay and elastic silts would be considered subject to cyclic softening with a potential for reduction in shear strength rather than “classic” cyclically induced liquefaction associated with loose, saturated granular soils.

To evaluate the potential for soil liquefaction to occur at the site, Crawford used the simplified procedure outlined by Youd et al. (2001) 4 and guidelines/modifications consistent with liquefaction evaluation outlined in the Caltrans Geotechnical Manual (Liquefaction Evaluation Module, January 2020), the 1997 and 2013 boring data and laboratory test results, design groundwater at elev. 5.0 feet, a site-to-fault distance of 7.9 miles, Maximum Moment Magnitude (Mmax) of 7.3, and a Peak Ground Acceleration (PGA) of 0.60g. The seismic/fault parameters used in our analysis are based on the result of the site-specific response analysis completed by Strong Motions, Inc. Refer to Appendix V that includes the results of our liquefaction triggering analysis.

Based on the foregoing and our analysis of the 1997 and 2013 boring data, potentially liquefiable soils are present at this site. In general, we identify the potentially critical liquefiable soil zones (Factor of Safety < 1.0) in Table 5.

Table 5: Potentially Liquefiable Soil Zones/Layers

Boring ID Nearest Support

Potentially Liquefiable Soil Zones/Layers Liquefaction

Factor of Safety

Estimated Seismic-induced

Settlement (in)

Elevation (feet, NGVD29)

Layer Thickness

(feet) B-4 Abut-1 None N/A N/A N/A

RW-13-010 Bent-2 5.0 to -10.5 15.5 0.08 to 0.34 8.9

B-2 Bents-2 and -3

5.0 to -11.0 -41.0 to -55.0 -61.0 to -72.0

16.0 14.0 11.0

0.06 to 0.23

0.09 to 0.13

0.19 to 0.50

11.7

A-13-015 Bents-3 and -4 5.0 to -6.5 11.5 0.09 to 0.27 3.9

B-1 Bent-4 5.0 to -2.0 7.0 0.19 2.2

B-7 Bent-6 -11.0 to -18.0 -51.0 to -56.0

7.0 5.0

< 1.0 *

0.09 1.9

B-6 Bent-7 -14.5 to -19.5 -60.0 to -99.5

5.0 39.5

< 1.0 *

0.10 to 0.64 9.8

B-5 Bent-8

-7.5 to -15.5 -62.0 to -70.0 -79.5 to -88.0 -92.0 to -96.5

8.0 8.0 8.5 4.5

0.08 0.17

0.26 to 0.47 0.35

8.2

A-13-009 Bent-9 -1.0 to -31.0 -53.5 to -66.0

30.0 12.5

0.09 to 0.35

0.10 to 0.19 16.2

A-13-008 Bent-9 0.0 to -15.0 15.0 0.14 to 0.23 6.2

B-3 Abut-10 5.0 to -5.0 -10.0 to -15.0

10.0 5.0

0.15 to 0.34

0.12 5.0

4 Youd, T. L., et al, Liquefaction Resistance of Soils: Summary Report from the 1996 NCEER and 1998 NCEER/NSF Workshops on Evaluation of Liquefaction Resistance of Soils, Journal of Geotechnical and Geoenvironmental Engineering, Vol. 127, No. 10, October 2001, pp. 817-833.

Boring ID Nearest Support

Potentially Liquefiable Soil Zones/Layers Liquefaction

Factor of Safety

Estimated Seismic-induced

Settlement (in)

Elevation (feet, NGVD29)

Layer Thickness

(feet)

A-13-013 Abut-10 Anchor 5.0 to -4.5 9.5 0.16 to 0.23 3.4

Notes: * No SPT data available, based on estimated apparent density of granular layers.

The elevations shown in the table are approximate.

NA = Not Applicable.

Soft and very soft clays and silts (bay mud) in several test borings may not be “liquefiable” but are prone to excessive movement/displacement (cyclic softening) under loading and provide little lateral support for piles. Approximate soft to very soft cohesive zones are identified in the borings in Table 6.

Table 6: Soft and Very Soft Cohesive Zones/Layers

Boring ID Nearest Support

Elevation (feet, NGVD29)

Layer Thickness

(feet) RW-15-023 Abut-1 5.0 to -12.0 17.0

A-13-010 Bent-2 -10.5 to -32.5 22.0

B-2 Bents-2 and -3 -11.0 to -41.0 30.0

A-13-015 Bents-3 and -4 None NA

B-1 Bent-4 None NA

B-7 Bent-6 -18.0 to -51.0 33.0

B-6 Bent-7 -19.5 to -60.0 40.5

B-5 Bent-8 -15.5 to -62.0 46.5

A-13-009 Bent-9 -30.5 to -53.0 22.5

A-13-008 Bent-9 -15.0 to -43.0 28.0

B-3 Abut-10 -10.5 to -15.5 5.0

A-13-013 Abut-10 Anchor None NA

Notes: The elevations shown in the table are approximate.

NA = Not Applicable.

10.2.3 SEISMICALLY INDUCED SETTLEMENT

The potential for seismically induced settlement is considered high. Calculated settlement from the boring data indicates high variability across the site and high variable magnitudes of settlement.

During a seismic event, ground shaking can cause densification of granular soils that can result in settlement of the ground surface. For bridge foundation design, we consider the potentially liquefiable soil layers identified in Table 5 to be subjected to reduction to residual shear strength values and seismically induced (vertical) settlement under conditions of strong ground shaking form a design earthquake event.

Based on our analysis, potential post-liquefaction ground settlement across the length of the existing bridge ranges from 2-inches to as much as 16-inches (Boring A-13-009). Appendix V includes our calculation results for seismically induced ground settlement.

The methods outlined by Idriss and Boulanger (2008)5 were used to estimate post-liquefaction settlement. The method follows the approach developed by Ishihara and Yoshimine (1992) that relates volumetric strain, SPT (N1)60cs [i.e., (N1)60 values corrected for fines content], and the factor of safety against liquefaction (FSL) to estimate the post-liquefaction settlement of a liquefied layer.

Seismically induced settlement can cause drag loads/negative skin friction forces on deep foundation.

10.2.4 SEISMIC SLOPE INSTABILITY

Due to presence of weak “bay mud” soils and potentially liquefiable layers at this site, stability of the east bank was evaluated using the Slide Version 5.0 computer program developed by Rocscience using the Simplified Bishop Method of slope stability calculation to generate non-circular failure surface trials. Graphical output from these calculations are included in Appendix V. Slope geometry for each case is based on site topography provided by the City. The soil profile is based on results of our subsurface investigation. The soil profile is based on results of our subsurface investigation. The existing and proposed slope geometry and bridge supports have been modeled using the following cases:

• Case 1: Existing slope and bridge foundations under static conditions,

• Case 2: Existing slope and bridge foundations with pseudostatic analysis (seismic load) based on historic earthquake shaking,

• Case 3: Proposed slope and bridge foundations under static conditions, and

• Case 4: Proposed slope and bridge foundations with pseudostatic analysis (design seismic load).

Static condition soil properties are based on field and laboratory testing of recovered soils from borings performed at the site. No slope movement has been reported at the site and static conditions would therefore require a factor of safety greater than unity.

Pseudostatic analyses were preformed using a seismic coefficient set to 1/3rd the peak ground acceleration or the Caltrans maximum seismic coefficient of 0.2g as appropriate. The largest recent earthquake near the site was the 1989 Loma Prieta earthquake. As no slope failures or distortions were noted at the bridge location after the Loma Prieta earthquake, it was conservatively assumed that a factor of safety of unity was achieved (i.e., fully weakened and

5 Idriss, I. M. and Boulanger, R. W., "Soil Liquefaction During Earthquakes," Earthquake Engineering Research Institute, pages 140-142 and 152-158, 2008.

pseudostable). This earthquake was estimated to have produced peak ground acceleration of approximately 0.3g at the site (0.1 seismic coefficient).

For pseudostatic analysis potentially sensitive soft soils were assumed to weaken during shaking. The reduction in strength used for the final model was determined by an iterative approach. The cohesion of Unit 2 silty clays was reduced until a factor of safety against sliding of approximate unity was achieved using a seismic coefficient of 0.1g (1989 Loma Prieta earthquake).