GM15-14_Dene_Tene_Path-Eska_Creek_Liquefaction_Analysis_Memo_29July2014.pdf

Memorandum

Western Federal Lands Highway Division 610 E. Fifth Street

Vancouver, WA 98661-3801

Date: July 29, 2014 In Reply Refer To: HFL-17

From: Diann Morehouse, P.E.

Geotechnical Engineer

To: Neal Christensen, P.E.

Project Manager

Subject: GEOTECHNIAL MEMORANDUM GM15-14 Liquefaction Analyses Dene Tene Path, Eska Creek Bridge

AK TTP CHIC 2013(1)

INTRODUCTION

1.1 Scope and Purpose

WFLHD Geotechnical Memorandum GM05-14 Preliminary Foundation Recommendations, Dene Tene Path, Eska Creek Bridge, AK TTP CHIC 2013(1), dated February 5, 2014 provided general geologic site conditions, analysis, and preliminary foundation recommendations for the proposed Dene Tene Pedestrian Path Bridge across Eska Creek. This memorandum provides liquefaction and settlement analyses and conclusions for the proposed pedestrian bridge.

1.2 Previous Geotechnical Reports

In 1997, the Alaska Department of Transportation and Public Facilities completed a geotechnical investigation for the replacement of the Glenn Highway Bridge over Eska Creek, in Sutton, Alaska. The borings for this investigation were drilled in the vicinity of the current highway bridge abutments. The proposed pedestrian bridge will be located approximately 24 feet north of the current highway bridge. The results of that investigation are provided in the report by the Alaska Department of Transportation and Public Facilities, titled, Foundation Report Glenn Highway: MP 55-61 Moose Creek to Sutton, Eska Creek Bridge No. 542, dated September 1997.

2.0 SEISMIC DESIGN

The American Association of State Highway and Transportation Officials (AASHTO) criteria for seismic design of ordinary bridges, allows for damage to occur when the structure is subjected to 1000-year strong ground motion, but does not allow for collapse of one or more spans and loss of life. Bridges may suffer damage that requires repair or partial replacement.

AASHTO, however, requires bridges to withstand small, more frequent earthquakes without damage.

Dene Tene Pedestrian Path – Eska Creek Bridge Liquefaction Analysis

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

3.0 SEIMSIC HAZARD ANALYSIS

3.1 Regional Seismicity

The active faults in the area create conditions where seismic design issues may have a significant impact on structure design. The Maximum Credible Earthquake (MCE) for the closest fault, the Castle Mountain, is a magnitude 7.18 event approximately 7 miles from the bridge site. The United States Geologic Survey (USGS) National Seismic Deaggregations of Seismic Hazard for PGA shows seismic contributions from Shallow Random Sources M5 to M7.3 and Alaska 37-mile (60km) Deep Seismicity having average seismic magnitudes of 6.18 and 5.82, respectively.

The contribution from the Shallow Random Sources M5 to M7.3 are shown slightly closer to the site (approximately 2 miles closer), but has a lower magnitude (6.18) associated event and the Alaska 37-mile (60km) Deep Seismicity is further from the site at approximately 50 miles.

Table 1 provides a list of regional faults, including the approximate distance and direction, orientation relative to the project, slip rate, slip sense, dip direction, and age of each fault.

Table 1: Regional Faults

Name Approx. Distance & Direction (from project) Slip Rate Slip

Sense Dip Direction Age

Castle Mountain Fault (Talkeetna West Section) 7 miles North & West <0.2 mm/yr. SS Vertical, north <1,600,000

Castle Mountain Fault (Susitna Section) 12 miles West 1 - 5 mm/yr. SS Vertical, north <15,000

Denali Fault (West Muldrow-Alsek Section) 125 miles North >5 mm/yr. SS Vertical <15,000

Denali Fault (Central Muldrow-Alsek Section) 300 miles East 1 - 5 mm/yr. SS Vertical <15,000

Susitna Glacier Fault 125 miles Northwest 1 - 5 mm/yr. T Northwest <150

Totschunda Fault 225 miles East >5 mm/yr. SS Vertical <150

NOTE: SS – Strike Slip, T – Thrust

3.2 Seismic Site Class

Table 2 shows a summary of borings 96-1 and 96-2 standard penetration test (SPT) N-values and laboratory test results. Based on the subsurface profile at the bridge site, the weighted average SPT N-values for the top 100 feet of subsurface materials was determined to be 48 blows per foot in Boring 96-1 and 20 blows per foot in Boring 96-2. Plasticity Index (PI) values range between 6 and 21.

According to the site class definitions specified in Table 3.10.3.1-1 of ASSHTO-LRFD Bridge Design Specifications 6th Edition, 2012 (AASHTO-LRFD). Site Class D is defined as stiff soil with average Standard Penetration Test (SPT) blow count for the upper 100 feet of the soil profile between 15 and 50 blows per foot, moisture content less than 40%, and a PI value less than 20. All the soil samples tested, with the exception of two samples, met the moisture content and PI requirements. Boring 96-1, Sample 35.1 to 37.4 feet has a PI value of 21 and Sample

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

40.0 to 41.3 feet has a moisture content of 40.4%. These two values are considered on the borderline and both values considerably higher than the other values.

Table 2: Summary of Borings 96-1 and 96-2 N-Values and Laboratory Test Results

Boring Number Depth (ft.) N-Value (N1)60

Gravel

Sand

%Fines (<200)

Moisture LL PI

96-1

5* 29 48 - - - - - - 10* 22 32 - - - - - - 15* 53 70 - - - - - -

20.0 – 22.6 49 59 58 34 8 - - -

25.0 – 27.0 96 109 59 34 7 - - -

30* 244 107 - - - - - -

35.1 - 37.4 9 9 - 25 75 27.1 31 21

40.0 - 41.3 172 97 65 25 10 40.4 - -

45* 90 83 - - - - - -

49.9 - 51.8 198 88 10 31 59 - - -

55* 121 84 - - - - - -

60.0 - 61.4 Refusal 80 - 30 70 - - -

70* 75 56 - - - - - -

80.0 - 82.0 119 70 66 24 10 - - -

90* 100 70 - - - - - - 100 - 101.0 Refusal 67 57 31 12 - - -

96-2

5* 8 11 - - - - - -

9.8 – 12.5 8 10 6 32 62 16.4 28 7

15. 0 - 17.0 77 78 67 27 6 - - -

20.0 - 22.6 6 5 6 76 18 18.5 27 6

25* 19 16 - - - - - -

29.9 - 30.8** 10 8 50 41 9 12.3 27 7

29.6 - 31.5** 10 8 58 34 8 - - -

35* 15 8 58 - - - - - 40* Refusal 11 - - - - - -

45.0 - 46.9 51 72 - 34 7 - - - 50* 130 35 59 - - - - -

60.0 - 62.0 127 83 - 23 11 - - - 70* Refusal 77 66 - - - - -

80.1 - 82.0 99 58 - 25 8 - - - 90* Refusal 54 67 - - - - -

Notes: * Assumed depth from 1997 boring logs, **Laboratory data from 1997 report shows different depths and laboratory test results.

3.3 2007 AASHTO Uniform Seismic Spectrum Accelerations Software

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

The Dene Tene Bridge site is located at 61.710594 degrees latitude and -148.899231 degrees longitude (NAD83). The AASHTO Seismic Design Parameters Software (version 2.10) was used to determine ground motion parameters for the site. The horizontal peak ground acceleration (PGA), short-period spectral acceleration (Ss), and long-period spectral acceleration (S1), for the referenced AASHTO weak rock site class (Site Class B) and a 1000-year return period (7 percent probability of exceedance in 75 years) and are shown in Table 3.

Corrected PGA and horizontal response spectral coefficients for Site Class D are distinguished as Scaled Spectrum Acceleration. Accelerations scaled for Site Class D, are shown in Table 4.

Table 3: AASHTO Uniform Spectrum Accelerations for 7% PE in 75 Years (Return Period 1000 years)

Period Period (seconds)

Acceleration (g)

Site Class

Seismic Zone

PGA 0 0.571 B 4

SS 0.2 1.294

S1 1 0.503

PGA - Horizontal Peak Ground Acceleration at Ss - Horizontal Response Spectral Acceleration at Period of 0.2 sec on Site Class B Rock S1 - Horizontal Response Spectral Acceleration at Period of 1.0 sec on Site Class B Rock

Table 4: AASHTO Scaled Spectrum Accelerations for Site Class D, 7% PE in 75 Years (Return Period 1000 years)

Periods Period (seconds)

Acceleration (g)

Site Class

Seismic Zone

AS 0 0.571 D 4

SDS 0.2 1.294

SD1 1 0.755

As - Scaled Peak Ground Acceleration for Site Class Listed SDS - Scaled Horizontal Response Spectral Acceleration at Period of 0.2 sec for Site Class Listed SD1 - Scaled Horizontal Response Spectral Acceleration at Period of 1.0 sec for Site Class Listed

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

3.4 Five-Percent Damped Design Response Spectrum

The seismic parameters corrected for Site Class D, Table 4, are used to develop the 5-Percent Damped Design Response Spectrum, shown in Figure A.

Figure A: 5% Damped Design Response Spectrum Corrected for Site Class D

3.5 Deaggregation

Deaggregation was performed for this project to determine the nearest sources contributing significantly to the seismic hazard at the bridge. The Deaggregation Hazard Data for a 5% in 50 years (975-year return period) is based on the USGS 1996 records was obtained from https://geohazards.usgs.gov/deaggint/1996/. The Deaggregation Hazard Data shows the principal sources of seismic hazard relative to the project site. A summary of Deaggregation Hazard Data for the project bridge site is shown in Table 5.

0.00

0.20

0.40

0.60

0.80

1.00

1.20

1.40

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.5 4.0

El as tic S ei sm ic C oe ffi ci en t, C sm

Period, T (sec)

5% DAMPED DESIGN RESPONSE SPECTRUM

https://geohazards.usgs.gov/deaggint/1996/

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

Table 5. Summary of Deaggregation Hazard Data for 975-Year Return Period.

Principal Sources % Contribution

Distance, R

(miles)

Moment Magnitude, M PGA Shallow Random Sources M5-M7.3 44.23 5.5 6.18

AK 37.2-mile deep seismicity 23.71 50 5.82

Castle Mountain Fault 10.21 7.7 7.18

1.0 Second PSA M9.2 Megathrust 11.8 51 9.2

1964-Zone M9.2 Megathrust 16.56 40 9.2

Shallow Random Sources M5-M7.3 35.27 7.76 6.64

Castle Mountain Fault 13.32 7.95 7.2

0.2 Second PSA Shallow Random Sources M5-M7.3 46.65 6.3 6.25

AK 37.2-mile deep seismicity 15.41 48 5.98

Castle Mountain Fault 11.27 7.3 7.15

3.6 Next Generation Attenuation Models Analysis

Earthquake ground motions at a specific site are influenced by fault characteristics such as distance from the fault planes, directivity, fault type, closest rupture distance, and dip angle. The Next Generation Attenuation (NGA) models were used to estimate weighted average peak ground acceleration based on the known fault characteristics. The mean plus one standard deviation of the five NGA relationships was determined for rock and the amplification factor from AASHTO (2013) based on the site class was used to determine the PGA at the ground surface for input into the liquefaction analysis. The Castle Mountain Fault was chosen for the attenuation analysis due to the close proximity to the project and the elevated moment magnitude value. Based on the average of five attenuation models, the PGA is 0.378 g.

3.7 Scaled Peak Ground Acceleration, AS

The Scaled Peak Ground Acceleration, AS, was re-calculated using the PGA of 0.378 g, which was determined by the NGA model, within the 5-Percent Damped Design Response Spectrum relationship. The Scaled Peak Ground Acceleration, AS for Site Class D is determined to be

0.424 g.

4.0 LIQUEFACTION

4.1. Liquefaction

Soil liquefaction is a phenomenon in which the effective stress in a cohesionless soil deposit below the groundwater drops to zero due to pore pressure generation resulting from earthquake shaking. Recently deposited and relatively loose natural soils and un-compacted or poorly compacted fill soils are susceptible to liquefaction. Submerged loose sand and silty sands are particularly susceptible to liquefaction. Loose silts and gravels may also have a potential for liquefaction. Dense natural, well-compacted fills and cohesive soils have a relatively low susceptibility to liquefaction.

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

4.2. Liquefaction Potential

Initial screening of site conditions for the potential for liquefaction was conducted to determine if a more detailed evaluation of liquefaction would be necessary. An evaluation for the potential for, and consequences of, liquefaction in soils near the surface were made in accordance with the following criteria:

• Ground water level is within 50 feet of the ground surface

• Seismic Hazard Level of the bridge, which depend on the value of Site Class-adjusted spectral acceleration at both long period (SD1) and short period (SDS)

• Geologic age and depositional environment are indicators of liquefiable materials.

Relatively young deposits (Holocene <11,000 years) deposited in potentially loose, uncompacted environments are susceptible to liquefaction.

• Site history can indicate if any liquefaction has occurred in the past.

• If soils are below the groundwater table, they may be susceptible to liquefaction.

• Clayey soils with low sensitivity have a potential for liquefaction if one or more of the following criteria are met:

• Clay content (grain size < 0.005 mm) less than 15 percent;

• Liquid limit less than 35 percent; or,

• Natural water content greater than 90 percent of the liquid limit.

• Clayey soils with high sensitivity should not be screened out. Sensitive clays possess all of the following properties:

• Liquid limit less than 40 percent.

• Water content greater than 0.9 times the liquid limit

• Liquidity index greater than 0.6

• (N1)60 less than five or normalized cone penetration resistance qc1, less than 20 ksf.

• Low plasticity silts and sands have a potential for liquefaction if:

• Minimum corrected Standard Penetration Test resistance, (N1)60 values are less than

25 blows per foot,

• Corrected cone penetration test tip resistance qc1, is less than 150 ksf with a sufficient number of tests,

• The normalized shear wave velocity, Vs1, is less than 660 ft./sec., or

• Or a geologic unit is present at the site that has been observed to liquefy in past earthquakes

4.3 Liquefaction Analysis Results

A consequence of liquefaction is the subsequent dissipation of excess pore-water pressure, leading to settlement. A liquefaction analysis was completed using the software program LiquefyPro, Version 5.8n, developed by CivilTech. Liquefaction was analyzed using Idriss &

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

Seed Fines Correction method. Settlement was analyzed using both Tokimatsu & Seed with M Correction method and Ishihara & Yoshimine Method.

Results from the analyses are provided in Attachment A. (Note that the simplified method used by the LiquefyPro Software for liquefaction analysis has only been validated up to depths of about 75 feet.)

4.3.1 Boring 96-1 (East Abutment)

Boring 96-1 has a near surface layer just below the water table and a layer about 2-foot thick at a depth between 33 and 35 feet, are shown to be susceptible to liquefaction. The calculated settlement is 0.49 inch by Tokimatsu & Seed Method. Liquefaction analyses for this boring under the existing conditions are shown in Attachment A, Plates 1-A and 1-B.

A liquefaction analysis was performed for the east GRS abutment under post-construction conditions. The results from both analyzed methods show a 1-foot thick zone susceptible to liquefaction with negligible liquefaction-induced settlement. The results from both analyses are shown in Attachment A, Plates 2-A and 2-B.

4.3.2 Boring 96-2 (West Abutment)

B96-2 has a near surface layer and a layer from about 19 and 35 feet are shown to be susceptible to liquefaction. The highest calculated liquefaction-induced settlement is about 7 inches, by Ishihara & Yoshimine Method. Both liquefaction analyses for this boring under the existing conditions are shown in Attachment A, Plates 3-A and 3-B.

A liquefaction analysis was performed for the west GRS abutment under post-construction conditions. The results of these analyses show liquefaction susceptibility between about 5 and 7 feet, 19 and 22 feet, and 28 and 34 feet. The highest calculated liquefaction-induced settlement is about 2 inches by Ishihara & Yoshimine Method. The results for both analyses are shown in Attachment A, Plates 4-A and 4-B.

5.0 CONCLUSIONS

The results from the liquefaction analyses for the existing site conditions show the west abutment has the greatest risk of liquefaction and an estimated total and differential settlement of approximately 7 inches during a design earthquake but the risk reduces following construction of the GRS-IBS due to the additional effective stress below the new fill.

The construction of the abutment and wing-walls provides some liquefaction mitigation by loading the liquefiable soils below the abutment and wing-walls. This loading increases effective stresses, which will reduce the potential for liquefaction and liquefaction-induced settlement. Modeling the abutment and wing-walls, post construction, shows a reduced risk of liquefaction and settlement under both abutments. Results from the analyses of the west abutment show a lower risk associated with liquefaction and a reduced risk of liquefaction-induced settlement from 7 inches to 2 inches.

Geotechnical Memorandum No.GM15-14

AK TTP CHIC 2013(1)

The geosynthetics within the GRS abutment and wing-walls, internally stabilize the mass, creating a gravity structure that behaves as one unit. The GRS is a flexible structure, which tolerates deformations due to foundation soils and has a high resistance to seismic loading.

Soils with corrected SPT (N1)60 values greater than 25-30 are generally not considered susceptible to liquefaction. Gravels are also not usually considered susceptible to liquefaction unless they are capped by a layer that prevents drainage such as frozen soil. Due to the low temperatures during the winter months at this site, a relatively high PGA, and a potential of frozen soil layer, there is an increased risk of liquefaction. A much more detailed investigation, testing, and analysis program would be necessary to better refine the liquefaction hazard at this site due to frozen soil zones overlying gravel layers. Further analysis should be considered if the estimated magnitudes of total and differential liquefaction-induced settlement cannot be tolerated by the proposed structure following the design earthquake event.

6.0 REFERENCES

AASHTO, 2012, AASHTO LRFD Bridge Design Specifications, Sixth Edition (2012 with 2013 Interim Revisions) American Association of State Highway and Transportation Officials, Washington, D.C.

Adams, M., et al, (2011) Geosynthetic Reinforced Soil Integrated Bridge System, Interim Implementation Guide, Report No. FHWA-HRT-11-026, Federal Highway Administration, 169p.

Alaska Division of Geological & Geophysical Surveys, Quaternary Faults and Folds, Interactive Map, Alaska Department of Natural Resources, http://maps.dggs.alaska.gov/qff/#17900000:9100000:3, Accessed 28Jan2014

FHWA, 2008, Federal Lands Highway Project Development and Design Manual (PDDM).

Federal Lands Highway, 2006, Technical Guidance Manual (TGM), A Geotechnical Supplement to the FLH Project Development and Design Manual

Alaska Division of Geological & Geophysical Surveys, Quaternary Faults and Folds, Interactive Map, Alaska Department of Natural Resources, http://maps.dggs.alaska.gov/qff/#17900000:9100000:3, Accessed 28Jan2014

ATTACHMENT A: Liquefaction Analysis http://maps.dggs.alaska.gov/qff/#17900000:9100000:3 http://maps.dggs.alaska.gov/qff/#17900000:9100000:3

Attachment A – Liquefaction Analysis