Ninole_Geotechnical_Report_Final.pdf

Final Geotechnical Report (updated February 25, 2016)

Ninole Bridge, Mamalahoa Highway, Route 11, District of Kau, Island of Hawaii, State of Hawaii

Prepared for

Federal Highway Administration Central Federal Lands Highway Division

12300 West Dakota Avenue Lakewood, CO 80228

December 15, 2015

Hirata & Associates, Inc.

99-1433 Koaha Place

Aiea, HI 96701

9191 South Jamaica Street

Englewood, CO 80112-5946

W.O. 15-5678.1 PAGE ii

Contents Section Page

1.0 Introduction

1.1 Background and Location

1.2 Purpose and Scope

2.0 Geology and Seismicity

2.1 Regional Geology

2.2 Seismic Design Parameters

3.0 General Subsurface Characterization

4.0 Analyses and Recommendations

4.1 Bridge Foundation

4.1.1 Bridge Foundation Bearing Resistance

4.1.2 Bridge Foundation Lateral Resistance

4.1.3 Probing and Grouting

4.2 Bridge Abutment Walls and Wing Walls

4.2.1 Active / At‐rest Pressures

4.2.2 Dynamic Earth Pressure

4.2.3 Drainage

4.2.4 Geosynthetic Reinforced Soil Backfill

4.3 Pavements

4.4 Site Grading

4.4.1 Site Preparation

4.4.2 Structural Excavations

4.4.3 Slope Gradients

4.4.4 Onsite Fill Material

4.4.5 Imported Fill Material

4.4.6 Compaction

5.0 Special Contract Requirements

6.0 Additional Services

7.0 Limitations

Appendixes

A Field Exploration Program & Boring Logs

B Laboratory Test Results

C Slope Stability Diagram

HIRATA & ASSOCIATES, INC.

W.O. 15-5678.1 PAGE 1

1.0 Introduction

1.1 Background and Location

This report provides geotechnical recommendations for the Ninole Bridge replacement located along Mamalahoa Highway, Route 11, in the District of Kau, Hawaii. Ninole Bridge is located at MP 56.7 and approximately 480 feet south of the Punaluu Road and Mamalahoa Highway intersection. Project vicinity and bridge photos are presented on Figures 1 and 2.

FIGURE 1

Project Vicinity Map

The Central Federal Lands Highway Division (CFLHD) of the Federal Highway Administration (FHWA), in cooperation with the Hawaii Department of Transportation is proposing to replace the existing deficient approximately 60‐feet long, 27.75‐feet wide, three span timber bridge over Ninole Stream to meet current design standards for roadway width, load capacity, bridge railing and transitions, and bridge approaches.

We understand that Option 1 presented in the Structure Selection Memo, dated March 2015, will be selected for replacement of the existing bridge. Option 1 includes a new 42.33‐feet wide, 65‐feet long single span bridge with wing walls that cantilever off the abutments. Geosynthetic reinforced soil will be placed behind the abutments. The abutments will be embedded into weathered basalt, and as a result, scour protection may be required for the abutments.

W.O. 15-5678.1 PAGE 2

FIGURE 2

Existing Bridge

1.2 Purpose and Scope

This report provides geotechnical recommendations based on engineering analyses of our subsurface investigation, including the general subsurface characterization, seismicity, boring logs, and laboratory test results. Geotechnical recommendations for the design of bridge abutments, abutment walls and wing walls, and site grading are presented in this report.

W.O. 15-5678.1 PAGE 3

2.0 Geology and Seismicity

2.1 Regional Geology

The project site is located in the Kau area on the south side of the Island of Hawaii. The Kau area is located in the Mauna Loa geomorphic province of Hawaii.

The island of Hawaii was built by the lava poured from five volcanoes, each with independent rift zones and individual geological history. The area of each volcano dome is partly dependent on its age. The southern slope of Kohala Mountain has been buried beneath Mauna Kea lavas. The western and southern slopes of Hualalai, and northwestern slopes of Kilauea have been buried by Mauna Loa lavas. The lavas of Mauna Loa interfinger with the lavas of Kilauea, Hualalai, and Mauna Kea. The lavas of Mauna Kea interfinger with the latest lavas of Kohala Mountain.

Hawaii Island, with the exception of the windward slope of Kohala, is little eroded. The only perennial streams are on the northeastern slopes of Mauna Kea and Kohala. The high permeability of the fresh lavas forming the surface of Kilauea, Mauna Loa, and Hualalai inhibit the development of permanent streams.

Large areas on these mountains are covered with black rock and are bare and devoid of vegetation. The southwestern side of Kilauea is a desert.

Mauna Loa is a shield‐shaped dome about 60 miles long and 30 miles wide. It is one of the most prolific producers of lava on earth. Few cinder cones lie on its slopes and within the summit caldera, Mokuaweoweo.

The project site is located in the Kau district on the southeastern slope of Mauna Loa. The area is blanketed with Pahala ash named after a village in the district. The ash crops out over about 450 square miles but most of it has been buried over by later lava flows. It is generally a fine yellowish to brownish red vitric ash forming an excellent soil and ranging from a few to 55 feet in thickness. There are also beds of lapilli and lithic ash derived from explosions at the Kilauea volcano.

The rocks of Mauna Loa are divided into three units of olivine basalt. The oldest is the Ninole volcanic series which forms the core of the mountain and is present in the partly‐filled canyons located on the southeast slope. The Ninole volcanic series is overlaid by Kau volcanic series and Pahala ash over a layer of Kahuku volcanic series.

2.2 Seismic Design Parameters

Based on the AASHTO site classification criteria, the predominant soils at the bridge site located at 19.13863o N latitude and ‐155.51732o W longitude can be classified as a rock profile corresponding to a Seismic Site Class B.

We understand that the 7th Edition of the AASHTO LRFD Bridge Design Specifications will be used for design of this project. Recommended soil parameters presented in this report are based on the 7th Edition of the AASHTO LRFD Bridge Design Specifications. Recommended seismic response parameters for use in design based on 7th Edition of the AASHTO LRFD Bridge Design Specifications represents ground motion corresponding to an exceedance probability of approximately 7 percent in 75 years for an earthquake with an approximate 1000‐year return period. The 1000‐year return period data for the soil at the bridge site was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal

W.O. 15-5678.1 PAGE 4

acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short and long‐period ground acceleration (SS and S1 respectively), and are summarized in Table 1.

TABLE 1

Summary of Seismic Parameters for Earthquake with 1000‐Year Return Period

Horizontal Peak Ground Acceleration, (PGA) 1.332g Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SS) 3.016g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (S1) 1.247g

Site Factor at Zero‐Period of Acceleration Spectrum, (Fpga) 1.0 Site Factor at Short‐Period Range of Acceleration Spectrum, (Fa) 1.0 Site Factor at Long‐Period Range of Acceleration Spectrum, (FV) 1.0

Factored Horizontal Peak Ground Acceleration, (AS) 1.332g Factored Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDS) 3.016g Factored Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 1.247g

Seismic Zone Zone 4

W.O. 15-5678.1 PAGE 5

3.0 General Subsurface Characterization

Table 2 presents a summary of the subsurface characterization. Boring logs and laboratory test results are presented in Appendixes A and B.

TABLE 2

Summary of Subsurface Materials for Foundation Design Purposes

Subsurface Depth (ft)

Material Description USCS Classification

Apparent Density

Ground Water Depth (ft)

BORING NN1 (ELEVATION +141.5) – NORTHEAST ABUTMENT, NINOLE BRIDGE

0.0 to 10.0 Silty GRAVEL GM Medium dense

GW

not encountered

10.0 to 17.5 BASALT WS‐WM Medium hard to hard

17.5 to 32.5 WEATHERED BASALT with clinker seams WM‐WH Medium hard to dense

32.5 to 40.5 BASALT WS Hard

BORING NN2 (ELEVATION +147.4) – SOUTHWEST ABUTMENT, NINOLE BRIDGE

0.0 to 5.5 GRAVEL GP‐GM Medium dense

GW

not encountered

5.5 to 17.0 BASALT WS Hard

17.0 to 31.0 WEATHERED BASALT with clinker seams WM‐WH Medium hard

31.0 to 35.0 BASALT WS Hard

W.O. 15-5678.1 PAGE 6

4.0 Analyses and Recommendations

This section provides a general interpretation of the site investigation and provides recommendations for design parameters at the proposed structures. Generalized subsurface profiles were developed based on data collected during the investigation. The primary structure considered for this project is replacing the existing timber structure with a longer single span concrete bridge.

4.1 Bridge Foundation

Based on as‐built plans dated November 1938, the existing bridge is founded on shallow foundations bearing directly on the basalt bedrock.

At the northeast abutment, plans indicate that the bottom of existing footing is approximately 12 feet below finish roadway elevation, at approximately +130.5. Boring NN1 encountered basalt at approximately 10 feet below existing roadway elevation at the rear of the abutment, at approximately +131.5.

At the southwest abutment, plans indicate that the bottom of existing footing is approximately 18 feet below finish roadway elevation, at approximately +129. Available rock profile plans indicate that the southwest abutment was constructed along or cut into a slope, with plans showing the basalt profile at a depth of about 7 feet at the rear of the abutment, at approximately +140. Boring NN2 drilled at the rear of the southwest abutment encountered basalt at approximately 5.5 feet, at approximately +142. Our borings generally confirm the rock profile shown on available plans.

Based on discussions with the Structural Engineer, we understand that the bottom of foundation elevation at the northeast and southwest abutments will be approximately +121.5 and +119.5, respectively. New foundations for the northeast and southwest abutments will be approximately 6.5 and 6 feet wide, and be embedded approximately 6.5 and 5.5 feet into weathered basalt, respectively. Based on our exploratory borings, medium hard to dense weathered basalt is anticipated at bottom of foundation elevations. As a result, shallow foundations bearing on or embedded into medium hard to dense weathered basalt are recommended for support of the replacement bridge.

4.1.1 Bridge Foundation Bearing Resistance

Bearing resistance factors are based on Table 10.5.5.2.2‐1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and the applied strength limit state resistance factor for bearing resistance for footings on rock is 0.45. A presumptive bearing value of 10,000 pounds per square foot at the Service Limit State was based on Table C10.6.2.6.1‐1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications and our engineering judgment. The anticipated settlement for the service limit state is less than 1‐inch.

Coring in weathered basalt resulted in RQD values ranging from 7 to 28 percent. Based on Section

10.6.3.2.1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, where engineering judgment does not verify the presence of competent rock, the competency of the rock mass should be verified using the procedures for Rock Mass Rating (RMR). A RMR range of 21 to 40, which represents a poor quality rock mass, was assumed for weathered basalt. Based on Table 10.4.6.4‐4 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, an estimated RMR of 23 was used for design of foundations bearing on weathered basalt.

Based on laboratory testing, unconfined compression tests on selected basalt core samples resulted in compressive strengths ranging from 3,518 to 8,107 pounds per square inch. Testing performed on basalt

W.O. 15-5678.1 PAGE 7

core samples from an adjacent project resulted in compressive strengths ranging from 1,615 to 4,135 pounds per square inch. An estimated compressive strength of 2,500 pounds per square inch was used for weathered basalt.

Semi‐empirical methods, Carter and Kulhawy (1988) referenced in the 7th Edition of the AASHTO LRFD Bridge Design Specifications and NCHRP Report 651, LRFD Design and Construction of Shallow Foundations for Highway Bridge Structures was used to confirm the Extreme Event Limit State value for weathered basalt. Local practice generally limits the Extreme Event Limit State value to three times the Service Limit State value.

The following bearing values may be used for design of shallow foundations bearing on or embedded into medium hard to dense weathered basalt. The Structural Engineer should confirm that the bearing values do not exceed the nominal strength of concrete, 0.3f’c.

TABLE 3

Bearing Value for Shallow Foundations

Soil Type Extreme Event Limit State Strength Limit State Service Limit State Medium hard to dense weathered basalt 30,000 psf 13,500 psf 10,000 psf

Footings should be a minimum 5 feet in width, and be embedded a minimum 5 feet below finish adjacent grade. Footings located on or near the top of slopes should be embedded such that a minimum horizontal distance of 5 feet is maintained between the bottom edge of footing and slope face.

The bottom of footing excavations should be cleaned of loose material prior to placement of reinforcing steel and concrete.

4.1.2 Bridge Foundation Lateral Resistance

Resistance to lateral loading may be provided by friction acting at the base of foundations, and by passive earth pressure acting on portions of foundations embedded in weathered basalt. The resistance from adhesion along the vertical faces of foundations may be used in conjunction with the passive resistance.

Resistance from adhesion may not be used along the bottom of foundations.

4.1.2.1 Coefficient of Friction

The following coefficients of friction may be used with dead load forces for foundations bearing on or embedded in weathered basalt. Sliding resistance factors are based on Table 10.5.5.2.2‐1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and the applied strength limit state resistance factors for shear sliding resistance is 0.85. The coefficients of friction were based on an interface friction angle between mass concrete and clean sound rock (hard basalt) of δ=35 degrees, obtained from Table 3.11.5.3‐1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications. Based on our engineering judgment, a slightly reduced interface friction angle between mass concrete and weathered basalt of δ=33 degrees was assumed in our analyses.

TABLE 4

Coefficient of Friction

Soil Type Extreme Event LimitState Strength Limit State Medium hard to dense weathered basalt 0.65 0.55

W.O. 15-5678.1 PAGE 8

4.1.2.2 Passive Pressure

For passive pressure considerations, the following uniform pressures may be used for foundations socketed into weathered basalt. Local practice generally assumes a uniform lateral bearing resistance equal to one‐ third the vertical bearing capacity for basalt. Analyses based on methods developed by Hoek and Brown (1988) and Hoek (2002) which utilize the RMR to determine the shear strength of fracture rock masses and Section 3.11.5.6‐2, Unfactored Simplified Earth Pressure Distributions for Permanent Nongravity Cantilevered Walls with Discrete Vertical Wall Elements Embedded in Rock in the 7th Edition of the AASHTO LRFD Bridge Design Specifications generally confirm the passive pressures for foundation design.

Resistance from passive pressure should be neglected in the upper 12 inches, and where the embedded portions of the structure are within a horizontal distance of 5 feet from the slope face or existing abutments.

A relatively small lateral movement of foundations into the weathered basalt, generally less than 0.01H where H is the height of the wall, is required to reach the maximum passive pressure. The passive pressures presented in Table 5 for foundations embedded into weathered basalt assume that recommendations presented in Section 4.1.3 Probing and Grouting will be implemented to confirm the presence of weathered basalt.

Sliding resistance factors are based on Table 10.5.5.2.2‐1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and the applied strength limit state resistance factors for passive pressure resistance to sliding is 0.5.

TABLE 5

Passive Pressure

Soil Type Extreme Event Limit State Strength Limit State Medium hard to dense weathered basalt 10,000 psf 5,000 psf

4.1.2.3 Adhesion for Side Shear Resistance

Additional resistance to lateral loads may be provided by side shear resistance between the foundation and weathered basalt. The following adhesion values for side shear resistance may be used provided that the footings are poured neat against weathered basalt.

The adhesion values were based on Horvath and Kenney (1979) for smooth rock sockets, with reduction factors for joints and fractures based on O’Neill and Reese (1999). Section C10.8.3.5.4b in the 7th Edition of the AASHTO LRFD Bridge Design Specifications indicates that methods to account for increased resistance due to adhesion due to sidewall roughness may be considered based on O’Neill and Reese (1999). Based on our engineering judgment of footings poured neat directly against basalt excavated with earth moving equipment resulting in non‐smooth sidewalls, and based on recommendations in FHWA‐NHI‐10‐016, Drilled Shafts: Construction Procedures and LRFD Design Methods, a regression coefficient of 1.0 in lieu of 0.65 was used for analyses.

A movement of 0.25‐inch is required for full side shear resistance to be mobilized.

Lateral resistance from adhesion along the bottom of foundations may not be combined with resistance from dead load forces and the coefficient of friction.

Sliding resistance factors are based on Table 10.5.5.2.4‐1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and the applied strength limit state resistance factors for side resistance in rock is 0.55.

W.O. 15-5678.1 PAGE 9

TABLE 6

Adhesion for Side Shear Resistance

Soil Type Extreme Event Limit State Strength Limit State Medium hard to dense weathered basalt 8,000 psf 4,400 psf

4.1.2.4 Lateral Load Resistance of Abutment Backfill

An abutment backfill stiffness of 4 kips per square foot per inch of deflection may be assumed for resistance to lateral loads in the longitudinal direction during a seismic event. Maximum lateral resistance of the abutment backfill should be limited to 5 kips per square foot.

4.1.3 Probing and Grouting

A probing and grouting program is recommended during construction due to occasional cavities/voids and clinker seams expected in the weathered basalt stratum. All footing excavations and areas adjacent to footings that utilize weathered basalt for resistance from passive pressure should be probed with a drill or air‐track hammer.

As a general guideline for cost estimating purposes, we recommend that one probe hole be drilled for every 50 square feet of foundation area for column footings, and at about 10 feet on centers along wall or strip footings. The probe holes should be a minimum 2 inches in diameter and extend to depths of about 15 feet below the bottom of footings. An approximate grout volume of 0.1 cubic yards per probe hole may be used for cost estimate purposes only, for the 2‐inch diameter, 15‐feet deep probe holes. Should voids be encountered in the weathered basalt stratum, the approximate grout volume may vary by magnitudes ranging between 10 and 50.

In areas adjacent to footings that utilize weathered basalt for resistance from passive pressure, we recommend that one probe hole be drilled at about 10 feet on centers along the footing face, and be offset approximately 3 feet, measured laterally, from the footing face. The probe holes should be a minimum 2 inches in diameter and extend to depths of about 5 feet below the bottom of footings. Based on a footing embedment of approximately 6.5 feet, an approximate grout volume of 0.08 cubic yards per probe hole may be used for cost estimate purposes only, for the 2‐inch diameter, 11.5‐feet deep probe holes. Should voids be encountered in the weathered basalt stratum, the approximate grout volume may vary by magnitudes ranging between 10 and 50.

When available, the foundation plans should be provided to our office for review, and to provide further recommendations for the probing and grouting plan, including quantity and depths of probe holes.

All probe holes should be filled with pumped sand‐cement grout through a grout pipe starting at the bottom of the probe hole. The grout should conform to FP‐14 725.13.a.1 with modifications presented in 5.0 Special Contract Requirements section of this report, and should have a minimum compressive strength of 2,500 pounds per square inch. During grout operations, the grout pipe should be inserted to the bottom of the hole and should be pulled at a controlled rate that will keep the end of the grout pipe within the body of the grout. Placement of thin wall plastic pipes in probe holes may be necessary to prevent holes from caving.

Areas encountering cavities or voids that consume large quantities of grout may require additional probe holes. Voids exposed at the bottom of foundation excavations or at the subgrade level of concrete slab areas should be opened and filled with grout.

W.O. 15-5678.1 PAGE 10

4.2 Bridge Abutment Walls and Wing Walls

Based on an approximate finish roadway elevation of +143 at the northeast abutment, bottom of foundation at approximately +121.5, and a 6.5‐foot embedment, the bridge abutment wall will have a maximum height of about 15 feet. At the southwest abutment, based on an approximate finish roadway elevation of +147, bottom of foundation at approximately +119.5, and a 5.5‐foot embedment, the bridge abutment wall will have a maximum height of about 22 feet. The wing walls will extend approximately 20 feet into the bridge approaches. Bridge abutment wall and wing wall foundations may be designed using recommendations in 4.1 Bridge Foundation section of this report.

The backfill material properties estimated for lateral design are based on the results of our laboratory testing and engineering judgment based on local practices.

TABLE 7

Summary of Backfill Material Properties for Lateral Design

Soil Type Unit Weight (γ) Friction Angle (φ) Cohesion (c) Gravel & Silty gravel 120 pcf 34o 0 psf Granular structural fill 130 pcf 34o 0 psf

4.2.1 Active / At-rest Pressures

For active and at‐rest earth pressure considerations based on Coulomb earth pressure theories and local practices for intact weathered basalt and basalt, the following equivalent fluid pressures may be used for insitu soils and unreinforced backfills. Walls that can tolerate little or no movement should be design for at‐ rest earth pressures. A movement of 0.001H at the top of wall by rotation or lateral translation is required to reach the minimum active pressure based on a dense granular backfill. H is the height of the wall.

TABLE 8

Active / At‐rest Pressures

Soil Type Level Backfill Condition Restrained / At‐rest Condition Gravel & Silty gravel / Structural fill 33 psf/ft 57 psf/ft

Weathered basalt 30 psf/ft 40 pst/ft Basalt 25 psf/ft 35 psf/ft

4.2.2 Dynamic Earth Pressure

Dynamic lateral earth pressure considerations were based on the Mononobe‐Okabe Method. Based on our engineering judgment, an increased friction angle of 38 degrees was used for the granular structural fill and gravel backfills. A dynamic lateral force of 74H2 pounds per lineal foot of wall length may be assumed for seismic loading with a maximum peak ground acceleration of 1.332g for a 1,000‐year return period. The above dynamic lateral force may be used for level, unreinforced well‐compacted granular structural backfill conditions where walls are free to move laterally up to 1 to 2 inches or rotate in the event of an earthquake.

As a result, a horizontal acceleration component of 0.666g was used in the analysis. The vertical acceleration component was assumed to be 0. H is the height of the wall in feet. The dynamic lateral force may be assumed to act through the mid‐height of the wall.

4.2.3 Drainage

To prevent buildup of hydrostatic pressures, retaining walls should be well‐drained. The standard of practice consists of placing a minimum 12‐inch thick layer of free‐draining gravel at the back of the wall. The gravel should extend from the base of the wall, around subdrains and/or weepholes, and up to within 12 inches of finish grade.

W.O. 15-5678.1 PAGE 11

Alternatively, prefabricated drainage geocomposites, such as Miradrain or J‐drain, may be used in lieu of the free‐draining gravel. As with the free‐draining gravel, the drainage geocomposites should be placed at the back of the wall, be connected with the weepholes and/or subdrains (in accordance with manufacturers specifications), and extend to within 12 inches of finish grade. For freestanding walls, the drainage system should be covered by at least 12 inches of well‐graded, well‐compacted material.

4.2.4 Geosynthetic Reinforced Soil Backfill

Due to the relatively high seismic forces, and to reduce the potential dynamic earth pressures acting on bridge abutment walls and wing walls, geosynthetic reinforced soil (GRS) backfill may be considered. GRS backfill consists of closely spaced layers of geosynthetic reinforcement, typically spaced 12 inches or less, and compacted granular fill material. The geosynthetic reinforcement are similar to tieback elements that provide tensile resistance to driving forces, while the closely spaced nature of geosynthetic reinforcement creates a confined soil mass column which reduces potential lateral soil displacements. Design of the GRS backfill was based on methods presented in NCHRP Document 187 ‐ Seismic Design of Geosynthetic‐ Reinforced Soil Bridge Abutments with Modular Block Facing, FHWA‐HRT‐11‐026 ‐ Geosynthetic Reinforced Soil Integrated Bridge System, and Army TM 5‐818‐8 ‐ Engineering Use of Geotextiles.

Static and dynamic earth pressures acting on bridge abutment walls and wing walls with GRS backfill, in lieu of unreinforced backfill, should be negligible based on the following recommendations.

Although static and dynamic earth pressures from the GRS backfill acting on bridge abutment walls and wing walls should be negligible during a seismic event as abutment walls and backfill generally move in unison, the potential of permanent lateral soil displacements of the GRS backfill exists. Initial consideration was given to the use of expanded polystyrene geofoam (EPS) to further reduce the potential effects of lateral soil displacements of the GRS backfill from surcharging the abutment walls and wing walls during or after seismic events. However, based on discussions with the Structural Engineer, we understand that during a seismic event, the abutment will need to engage the backfill immediately to utilize resistance from the abutment backfill stiffness. As a result, EPS Geofoam was not included in the bridge abutment design.

4.2.4.1 Subgrade Preparation

Prior to installation of the GRS backfill, the excavation slope into the native gravel and silty gravel behind the GRS backfill should be cut and continuously benched. We recommend maximum bench heights of 4 feet, with minimum bench widths of 8 feet. The intent of the continuously benched native slope is to provide a horizontal surface to utilize frictional resistance to help resist the potential for sliding of the GRS backfill during a seismic event.

Slope stability analyses were performed on 15‐foot and 22‐foot high slopes for the northeast and southwest abutments, respectively, using Slope/W 2012. The slopes were continuously stepped‐up in 4‐foot high increments, with bench widths of 8 feet, from the toe to top of slope. A surcharge load was applied to the cut slope to simulate the weight of the backfill. For seismic (pseudo‐static) loading considerations, the continuously benched slope was assumed to accept some deformations. Therefore, for a peak ground acceleration of 1.332g, a horizontal seismic coefficient of 0.666g was utilized in our analysis. Analyses resulted in factors of safety of 1.53 and 1.87 for the northeast and southwest abutments, respectively. The minimum factor of safety for seismic loading conditions is 1.1. The slope profiles and critical slip surfaces are presented on Figures C‐1 and C‐2, Ninole Bridge Cut Backfill Slope on Plates C1.1 and C1.2.

W.O. 15-5678.1 PAGE 12

As a result, the continuously benched native slope should remain stable during a seismic event and the potential for a slope failure exerting significant forces on the GRS backfill will be reduced with the continuously benched native slope.

The bottom of excavation should extend to the bottom of abutment walls and wing walls, and should be scarified to a minimum depth of 6 inches, moisture conditioned to about 2 percent above optimum moisture content, and compacted to a minimum 95 percent compaction as determined by AASHTO T‐180 (ASTM D 1557). Where basalt or weathered basalt is encountered or exposed, scarification may be terminated prior to the minimum 6 inch depth.

If encountered, loose soils indicated by pumping conditions should be removed down to competent material and replaced with either approved onsite soil or imported fill compacted to a minimum 95 percent compaction as determined by AASHTO T‐180 (ASTM D 1557).

4.2.4.2 Geosynthetic Reinforcement

Based on FHWA‐HRT‐11‐026, a biaxial, woven polypropylene geotextile with a minimum ultimate strength of 4,800 pounds per foot and a minimum tensile strength at the 2 percent strain value of 1,320 pounds per foot is recommended. The geotextile should be vertically spaced at maximum 12‐inch intervals extending from the bottom of the bridge abutment wall to the bottom of the road section. The geotextile should conform to FP‐14 714.04, Type IV or greater. The layers of geotextile should extend laterally from the rear of the bridge abutment walls to the cut and continuously benched native slope face.

Our analyses were based on 15‐foot and 22‐foot high abutments and minimum geotextile lengths of 16 feet and 20 feet for the northeast and southwest abutments, respectively, extending from the rear of the bridge abutment wall to the bottom of the cut and continuously benched native slope face. The minimum geotextile length will increase by the bench width, 8 feet, at each bench interval. External stability analyses included sliding and overturning of the GRS backfill for both static and dynamic conditions, and analyses resulted in acceptable factors of safety of greater than 1.5. Internal stability analyses included pullout and breakage of the geotextile for both static and dynamic conditions, and analyses resulted in the allowable reinforcement strength and the minimum tensile strength at the 2 percent strain value greater than the required reinforcement strength.

To prevent soil raveling between the reinforcement elements, an overlapping length of 3 feet may be used to wrap the compacted soil layers around the face of the backfill and embedded into the backfill for the northeast abutment. For the southwest abutment, an overlapping length of 5 feet for the upper most reinforcing element, and 3 feet for the remaining lower elements may be used to wrap the compacted soil layers around the face of the backfill and embedded into the backfill.

4.2.4.3 GRS Fill Material

Aggregate base course is recommended for the reinforced fill material. The aggregate base course should conform to FP‐14 703.05, Gradation Designation C (Base). The aggregate base course should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 95 percent compaction as determined by AASHTO T‐180 (ASTM D 1557).

4.3 Pavements

The AC concrete pavement encountered in borings NN1 and NN2 consisted of approximately 4 and 3.5 inches of asphaltic concrete, respectively. Approximately 4.5 inches of base material was observed in boring

W.O. 15-5678.1 PAGE 13

NN2, while base material was not observed or could not be distinguished from the underlying subgrade in boring NN1.

Due to the relatively short length of pavement for bridge approaches, based on discussions with Central Federal Lands Highway Division, we understand that new pavement will generally match the existing pavement sections identified in the borings.

4.4 Site Grading

4.4.1 Site Preparation

The project site should be cleared of all vegetation, including large tree roots, boulders, asphaltic concrete pavement, existing footings, and other deleterious material. Prior to placement of fill, the exposed subgrade should be scarified to a minimum depth of 6 inches, moisture conditioned to slightly above the optimum moisture content, and compacted to a minimum 95 percent compaction as determined by AASHTO T‐180 (ASTM D 1557). Where basalt or weathered basalt is encountered or exposed, scarification may be terminated prior to the minimum 6 inch depth.

4.4.2 Structural Excavations

Based on our exploratory borings, we believe that excavations into the onsite gravel and silty gravel can generally be accomplished using conventional excavating equipment. Excavations into medium hard to hard basalt and weathered basalt will require pneumatic equipment.

Temporary cuts into the near surface gravel and silty gravel should be stable at slope gradients of 1H:1V or flatter. Temporary cuts into the basalt and weathered basalt should stand at a near vertical gradient.

However, it should be the Contractor’s responsibility to conform to all OSHA safety standards for excavations.

4.4.3 Slope Gradients

Permanent cut and fill slopes may be designed for the following gradients.

• Permanent fill slopes 2H:1V or flatter

• Cuts in gravel and silty gravel 2H:1V or flatter

• Cuts in basalt and weathered basalt 1H:1V or flatter

4.4.4 Onsite Fill Material

The onsite gravel and silty gravel will be acceptable for reuse in structural fills and backfills provided all rock fragments larger than 3 inches in maximum dimension are removed prior to reuse. Excavated basalt and weathered basalt will also be acceptable for reuse in structural fills and backfills provided the material is crushed to a well‐graded consistency with a maximum particle size of 3 inches.

4.4.5 Imported Fill Material

Imported structural fill should be well‐graded, non‐expansive granular material and conform to FP‐14 704.04 with modifications presented in 5.0 Special Contract Requirements section of this report.

W.O. 15-5678.1 PAGE 14

4.4.6 Compaction

Structural fill and backfill should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 95 percent compaction as determined by AASHTO T‐180 (ASTM D 1557).

Fill placed in areas which slope steeper than 5H:1V should be continually benched as the fill is brought up in lifts. Fill placed on slopes should be keyed and benched into the existing slope to provide stability for the new fill against sliding. Filling the slope with sliver fills should be avoided.

W.O. 15-5678.1 PAGE 15

5.0 Special Contract Requirements

Section 261. – Reinforced Soil Slopes and Geosynthetic Reinforced Soil (GRS) Backfill

Description

261.01 This work consists of constructing reinforced soil slopes and GRS backfill.

Material

261.02 Conform to the following Subsections:

Reinforced Soil Slope Backfill material 704.09 GRS Backfill material 703.05 (Gradation C) Reinforcement geotextile and geogrid 714.04

Construction Requirements

261.03 Submittals. At least 30 days before starting reinforced soil slope and GRS backfill construction, submit the following according to Subsection 104.03:

(a) Manufacturer’s certification. Certification and test results indicating the proposed soil reinforcement, reinforced embankment, and facing material meets requirements;

(b) Splice details. Detail proposed splices in soil reinforcement; and

(c) Facing details. Provide installation methods and connection details for facing system.

261.04 General. See Section 257 for Contractor‐designed reinforced soil slope systems.

Verify the limits of the reinforced soil slope and GRS backfill installation.

Clear the area of vegetation and obstructions according to Sections 201 and 203.

Conserve topsoil according to Subsection 204.05.

Excavate into the gravels and bedrock using mechanical means. Blasting is not allowed.

Ensure that the bottom of the excavation is level and even.

Prepare and compact foundation soils according to Subsection 204.09(a). Where basalt is encountered or exposed, scarification may be terminated prior to the minimum 6 inch depth. Grade the foundation soils supporting the reinforced soil slope and GRS backfill for a width equal to the length of the lowest soil reinforcement level. Bench cut slope surfaces in 4‐foot high increments with bench widths of 8 feet to key the reinforced soil slope and GRS backfill into the existing embankment.

261.05 Soil Reinforcement and Facing. Place the bottom layer of reinforcement so that it is in full contact with the level subgrade and so that it extends from the rear of the abutment or slope face to the limits shown on the plan drawings. Ensure that no voids exist beneath the reinforcement. Place the reinforcement so that the strongest direction is perpendicular to the abutment of slope face.

W.O. 15-5678.1 PAGE 16

Lay each layer of soil reinforcement flat, Pull the reinforcement tight, and hold in place with pins, soil piles, or other approved methods. Construct the reinforced soil slope and GRS backfill to achieve the slopes specified and roadbed tolerances according to Subsection 204.13(d).

Wrap the reinforcement as shown on the plan drawings. When seams are necessary, stagger at least 24‐ inches apart on subsequent lifts. Seams are only allowed in the direction perpendicular to the abutment or slope face. Seams of adjacent sheets of reinforcement shall overlap a minimum of 12 inches. Except at the wrapped face of the slope or abutment, an approximately 1‐inch thick layer of backfill shall be placed between the two overlapping reinforcement sheets.

Grade and compact backfill material according to Subsections 204.09 and 204.11 in maximum 8‐inch loose lifts before placing the next soil reinforcement layer. No equipment is allowed directly on the reinforcement. Maintain a minimum cover of 6 inches (150 millimeters) over the reinforcement during spreading and compacting of fill material. Avoid sudden stops, starts, or turns of the equipment. Do not use sheepsfoot‐rollers. Use lightweight mechanical tampers, rollers, vibratory systems, or other methods for compaction within 36 inches (900 millimeters) of the slope face.

Place backfill material in a manner that tensions the reinforcement working from the slope face or abutment to the back of reinforcement. Remove any wrinkles that form during backfilling operations.

Do not damage or disturb the reinforcement elements. Do not wrinkle, tear, or displace the reinforcement. Carefully correct all damaged, misaligned, or distorted reinforcement elements.

Fill and compact any depressions caused during compaction so that the compacted lift is level prior to constructing the next layer.

Do not leave reinforcement exposed at end of shift. At the end of each shift, shape to drain and compact the work area to a uniform cross‐section. At the end of the day’s operation, slope the last lift of backfill away from the slope face to direct surface runoff away from the slope.

261.06 Acceptance. Material for reinforced soil slopes and GRS backfill will be evaluated under Subsections 106.02, 106.03, and 106.04.

Construction of reinforced soil slopes will be evaluated under Subsections 106.02 and 106.04.

Excavation will be evaluated under Section 204.

Measurement

261.07 Measure the Section 261 pay items listed in the bid schedule according to Subsection 109.02.

When measuring reinforced soil slope and GRS backfill within the stabilized volume by the cubic yard (cubic meter), measure in place.

Payment

261.08 The accepted quantities will be paid at the contract price per unit of measurement for the Section 261 pay items listed in the bid schedule. Payment will be full compensation for the work prescribed in this Section. See Subsection 109.05.

W.O. 15-5678.1 PAGE 17

Section 704.04 Structural Backfill. Furnish a suitable, well‐graded, non‐expansive granular material conforming to the following:

(a) Maximum particle size 3 in (75 mm)

(b) Material passing No. 200 (75‐μm) between 8 and 20 percent sieve, AASHTO T 27 and AASHTO T 11

(c) Plasticity index, AASHTO T 90 10 max.

(d) California Bearing Ratio, AASHTO T 193 CBR value, 15 min.

(ASTM D 1883) CBR expansion value, 1.0 percent max.

Section 725.13 Grout.

(a) Neat hydraulic cement grout.

(1) Ground anchors, micropiles, soil nails, and probe holes. Furnish a pumpable mixture of portland cement, sand, water, and admixtures. Use Type II portland cement conforming to AASHTO M 85. Furnish sand according to Subsection 703.01. Furnish water according to Subsection 725.01(a). Provide the minimum water content necessary for placement and a maximum water/cementitious material ratio between 0.40 and 0.50 by mass.

Chemical admixtures that control bleed or retard set may be used provided the additives conform to Subsection 711.03 and are mixed according to the manufacturer’s recommendations. Do not use admixtures containing more than trace (from impurities, not as an intended constituent) quantities of chlorides, fluorides, aluminum, zinc, or nitrates.

The maximum water‐soluble chloride ion (Cl‐) content of the grout is 0.06 percent CL‐ by mass of cement when tested according to ASTM C1218.

Furnish grout conforming to the following minimum compressive strengths:

(a) Ground anchors 3,500 lb/in2 (25 MPa) at 7 days

(b) Micropiles 2,500 lb/in2 (17.5 MPa) at 7 days and 5,000 lb/in2 (35MPa) at 28 days

(c) Soil nails 1,500 lb/in2 (10.5 MPa) in 3 days and 3,000 lb/in2 (21 MPa) in 28 days

(d) Probe holes 2,500 lb/in2 (17.5 MPa) at 28 days

W.O. 15-5678.1 PAGE 18

6.0 Additional Services

We recommend that we perform a general review of the final design plans and specifications. This will allow us to verify that the foundation design and earthwork recommendations have been properly interpreted and implemented in the design plans and construction specifications.

For continuity, we recommend that we be retained during construction to (1) observe all probing and grouting operations, (2) observe footing excavations prior to placement of reinforcing steel and concrete, (3) review and/or perform laboratory testing on import borrow to determine its acceptability for use in compacted fills, (4) observe structural fill placement and perform compaction testing, and (5) provide geotechnical consultation as required.

Our services during construction will allow us to verify that our recommendations are properly interpreted and included in construction, and if necessary, to make modifications to those recommendations, thereby reducing construction delays in the event subsurface conditions differ from those anticipated.

Appendix A Field Exploration Program & Boring Logs

W.O. 15-5678.1 PLATE A1.1

Appendix A: Field Exploration Program & Boring Logs

A1.0 Introduction Hirata & Associates, Inc. completed an initial field exploration program for Ninole Bridge, Mamalahoa Highway, Route 11, District of Kau, Island of Hawaii, State of Hawaii, Task Number 499079.08.SU.HI, between September 4 through 10, 2014. The scope of work for the initial field exploration program included drilling a total of 2 borings. The scope of work included 2 additional borings, which was planned to be drilled after the final detour bridge location and span were determined. However, at the time of this report, Central Federal Lands Highway Division decided not to proceed with the additional exploratory boring. Each component of the field exploration program was coordinated and observed by a Geotechnical Engineer from Hirata & Associates, Inc. Boring location plans, as well as individual boring logs are attached.

These logs represent a compilation of field data and description of the soil samples by Hirata & Associates, Inc. personnel. The methods used to conduct the field exploration program are described below.

Representative soil and rock samples collected during the field exploration program were transported to Hirata & Associates, Inc. in Aiea, Oahu, Hawaii for testing. A summary of the laboratory testing program is provided in Appendix B.

A2.0 Explorations Hirata & Associates, Inc. of Aiea, Oahu, Hawaii provided the drilling services for the soil borings. Borings were completed using a Mobile CME‐55 truck‐mounted drill rig.

The borings were drilled between September 4 through 10, 2014. Two borings at the rear of existing bridge abutments, NN1 and NN2, were completed to depths of about 35 and 40.5 feet below ground surface to explore subgrade conditions. Borings were drilled and sampled using continuous flight solid stem augers and continuous sampling rock coring techniques.

The locations of individual borings were located in the field by measuring/taping offsets from existing features shown on the plans. Boring locations are shown on the Boring Location Plan, Plate A2.2, and are shown below in Table A‐1. Surface elevations at boring locations were estimated based on the Topographic Survey provided by CH2M Hill received on October 31, 2014. Following drilling activities, the borings were backfilled with cuttings generated during the drilling operations and capped with “Instant Road Repair” by International Roadway Research in accordance with applicable local, state, and federal regulations.

TABLE A‐1

Field Exploration Locations

Foundation Designation

Boring Number Station Offset

Ground Elevation

Termination Elevation

Basalt Elevation

NE Abutment NN1 850+33 5 LT 141.5 ft 104 ft 131.5 ft SW Abutment NN2 849+54 5 RT 147.4 ft 112.4 ft 141.9 ft

A3.0 Soil and Rock Sampling Disturbed samples were obtained from the borings in general accordance with ASTM D 1586 for the Standard Penetration Test (SPT). The SPT involves driving a 2‐inch outside diameter, 1.375‐inch inside diameter split‐spoon sampler a depth of 18 or 24 inches with a 140‐pound hammer falling a distance of 30 inches. The number of blows required to advance the split‐spoon sampler through each of the 6‐inch

W.O. 15-5678.1 PLATE A1.2

increments was recorded. The SPT resistance, or N‐value, is defined as the number of blows required to drive the sampler over the second and third 6‐inch increments. The N‐value provides a means for evaluating the relative density or compactness of cohesionless (granular) soils and consistency or stiffness of cohesive (fine‐grained) soils. Because high penetration resistance prevented driving the total length of the sampler at times, the penetration resistance for the partial penetration was recorded where applicable. Representative portions of the split‐spoon sample obtained in conjunction with the SPT were placed in plastic bags and transported to the Hirata & Associates, Inc. for testing.

Samples were also obtained using a modified California (MC) barrel sampler in general accordance with ASTM D 3550. The MC test procedure is similar to the SPT, except a large diameter barrel sampler (3‐inch O.D., lined with 2.42‐inch‐diamter brass sleeves) is used and driven 18 inches. As a result of the larger diameter, the MC sampler yields slightly higher raw blow counts when compared to SPT N‐values for similar soils. However, because the blow counts differ slightly, and because the difference in blow count does not significantly impact our results, the raw MC blow counts over the second and third 6‐inch increments were used to define the relative density and consistency of the subsurface materials. MC samples were sealed with plastic end caps and transported to Hirata & Associates, Inc. for testing.

A bulk soil sample was recovered from boring NN2 between depths of about 1 to 3 feet below ground surface. The bulk soil sample was placed in a plastic bag and transported to Hirata & Associates, Inc. for testing.

Rock samples were collected from borings using NX and HQ core barrels having inside diameters of 2.1 and

2.5 inches, respectively. During coring, cuttings were removed from the borehole by circulating water down the drill rods and back up the annulus of the boring. Core runs ranged between approximately 1.5 and 5 feet in length. Percent recovery and rock quality designation (RQD) values are shown on the boring logs.

Percent recovery is the ratio of the length of recovered core to the total length of the core run. RQD is the ratio of the sum of the length of intact core pieces greater than 4 inches long in a run to the total length of the core run. Following drilling, the cores were logged by Hirata & Associates, Inc. personnel, then stored in cardboard core boxes and transported to the Hirata & Associates, Inc. laboratory.

A4.0 Soil and Rock Classification System During the completion of borings, Hirata & Associates, Inc. personnel collected soil/rock samples and prepared field logs of the borings. Soil identification and descriptions, as shown on the field logs, are based on ASTM D2488, a systematic process for identifying and describing individual soil samples by visual and manual means. When sufficient laboratory testing was completed, select samples from borings were classified using the Unified Soil Classification System (USCS). Both the visual soil identification system and the referenced soil classification systems are summarized in the attached Boring Log Legend and USCS plates. Rock samples were classified based on the stratigraphic structure, rock strength, degree of weathering, and other properties. The rock classification system is summarized in the attached Rock Weathering Classification System.

Appendix B Laboratory Test Results

W.O. 15-5678.1 PLATE B1.1

Appendix B: Laboratory Test Results

B1.0 Introduction Laboratory tests were completed on select soil and rock samples recovered from the field exploration program in general accordance with ASTM testing methods.