[Supplemental]_Walker-Murray_Geotechnical_Report.pdf

GEOTECHNICAL AND BRIDGE FOUNDATION REPORT

Walker/Murray Improvements SW Walker Road and SW Murray Boulevard Intersection Washington County, Oregon

For Washington County Department of Land Use and Transportation c/o WH Pacific, Inc.

June 23, 2020

GeoDesign Project: WashCo-56-03

9450 SW Commerce Circle, Suite 300 l Wilsonville, OR 97070 l 503.968.8787 www.geodesigninc.com

June 23, 2020

Washington County Department of Land Use and Transportation c/o WH Pacific, Inc.

9755 Barnes Road, Suite 300 Portland, OR 97225

Attention: Wayne Bower, P.E.

Geotechnical and Bridge Foundation Report Walker/Murray Improvements

SW Walker Road and SW Murray Boulevard Intersection Washington County, Oregon

GeoDesign Project: WashCo-56-03

GeoDesign, Inc. is pleased to submit our geotechnical and bridge foundation report for the proposed the Walker/Murray Improvements in Washington County, Oregon. This report presents a description of our services and site conditions; a summary of our field explorations and laboratory testing; and our recommendations for design and construction of bridge foundations, embankment fills, and retaining structures. Our recommendations were developed in general accordance with OSSC and standards presented in AASHTO LRFD (2017).

We appreciate the opportunity to be of continued service to you. Please call if you have questions concerning this report or if we can provide additional services.

Sincerely, GeoDesign, Inc.

Nick Paveglio, P.E. Jeffery D. Tucker, P.E., G.E.

Senior Associate Engineer Principal Engineer cc: Andy Bowen – WH Pacific (via email only)

NNP:JDT:kt:sn

Attachments

One copy submitted (via email only)

Document ID: WashCo-56-03-062320-geor-rev.docx

© 2020 GeoDesign, Inc. All rights reserved.

i WashCo-56-03:062320

EXECUTIVE SUMMARY

Based on information provided by the design team and the results of our explorations, laboratory testing, and analyses, it is our opinion that the project is feasible from a geotechnical perspective. The primary geotechnical considerations for the project are summarized as follows:

The bridges should be supported on driven piles that extend into the very stiff silt and hard clay beneath the site. Driven piles will generate vibrations during installation. Based on experience and published information, we do not anticipate that structures within 50 feet of the piles will be damaged by vibrations.

The subgrade soil in the areas of the proposed culvert replacements is suitable to support the precast box culvert sections, provided weak and unsuitable subgrade materials are removed and replaced with structural fill and a 2-foot-thick layer of granular material is provided below the culverts.

The near-surface soil at the site is fine-grained and easily disturbed during wet weather. The contractor should protect the subgrade areas from disturbance during construction.

WashCo-56-03:062320

TABLE OF CONTENTS PAGE NO.

ACRONYMS AND ABBREVIATIONS

1.0 INTRODUCTION 1

2.0 PURPOSE AND SCOPE 1

3.0 SITE CONDITIONS 2

3.1 Geologic Setting 2

3.2 Surface Conditions 3

3.3 Subsurface Conditions 3

3.4 Infiltration Testing 4

4.0 LABORATORY TESTING 5

5.0 CONCLUSIONS 5

6.0 DESIGN 5

6.1 Permanent Slopes 5

6.2 Fills 5

6.3 Precast Box Culvert Recommendations 6

6.4 Seismic Considerations 6

6.5 Foundation Analysis and Design Recommendations 10

6.6 CIP Retaining Walls 14

6.7 Prefabricated Modular Block Walls 17

6.8 Groundwater Considerations 20

6.9 Site Drainage 20

7.0 CONSTRUCTION 20

7.1 Site Preparation 20

7.2 Erosion Control 22

7.3 Excavation 22

7.4 Materials 23

8.0 OBSERVATION OF CONSTRUCTION 26

9.0 LIMITATIONS 27

REFERENCES 29

FIGURES

Vicinity Map Figure 1

Site Plan Figure 2

TABLE OF CONTENTS PAGE NO.

APPENDICES

Appendix A Field Explorations A-1 Laboratory Testing A-1 Exploration Key Table A-1 Soil Classification System Table A-2 Boring Logs Figures A-1 – A-10 Atterberg Limits Test Results Figure A-11 Grain Size Testing Results Figure A-12 Summary of Laboratory Data Figure A-13 Appendix B Driven Pile Design Information B-1 Factored Axial Resistance – Murray Boulevard Bridge – Bent 1 Figure B-1 Factored Axial Resistance – Murray Boulevard Bridge – Bent 2 Figure B-2 Factored Axial Resistance – Walker Road Bridge – Bent 1 Figure B-3 Factored Axial Resistance – Walker Road Bridge – Bent 2 Figure B-4 Appendix C LPILE Parameters C-1 LPILE Soil Parameters – Murray Boulevard Bridge – Bent 1 Table C-1 LPILE Soil Parameters – Murray Boulevard Bridge – Bent 2 Table C-2 LPILE Soil Parameters – Walker Road Bridge – Bent 1 Table C-3 LPILE Soil Parameters – Walker Road Bridge – Bent 2 Table C-4 Appendix D Modular Block Retaining Wall Design Output D-1 Output Plots

ACRONYMS AND ABBREVIATIONS

AASHTO American Association of State Highway and Transportation Officials AC asphalt concrete ASTM American Society for Testing and Materials BGS below ground surface CIP cast-in-place CSZE Cascadia subduction zone earthquake FHWA Federal Highway Administration g gravitational acceleration (32.2 feet/second2) GDM Geotechnical Design Manual H:V horizontal to vertical LRFD Load and Resistance Factor Design MSE mechanically stabilized earth NA not applicable ODOT Oregon Department of Transportation OSHA Occupational Safety and Health Administration OSSC Oregon Standard Specifications for Construction (2018) pcf pounds per cubic foot pci pounds per cubic inch PDA Pile Driving Analyzer® PGA peak ground acceleration psf pounds per square foot psi pounds per square inch SPT standard penetration test USGS U.S. Geological Survey WEAP wave equation analysis program

1 WashCo-56-03:062320

1.0 INTRODUCTION

This report presents our geotechnical evaluation for the proposed Walker/Murray Improvements project in Beaverton, Oregon. The project consists of improvements to the intersection of SW Walker Road and SW Murray Boulevard, including replacement of the Walker Road and Murray Boulevard bridges over Cedar Mill Creek and the replacement of the culverts on North Johnson Creek under SW Far Vista Drive and SW Walker Road. The site location with respect to the surrounding physical features is shown on Figure 1. The existing topographic setting of the site is provided on Figure 2. Acronyms and abbreviations used herein are defined above, immediately following the Table of Contents.

The existing Murray Boulevard bridge over Cedar Mill Creek is located on SW Murray Boulevard approximately 275 feet south of the intersection of SW Murray Boulevard and SW Walker Road.

The bridge is approximately 66 feet long, 93 feet wide, and supported by driven piles. The existing Walker Road bridge over Cedar Mill Creek is located on SW Walker Road approximately 200 feet east of the intersection of SW Murray Boulevard and SW Walker Road. The bridge is approximately 66 feet long, 88 feet wide, and supported by driven piles.

The existing North Johnson Creek culvert on SW Walker Road is 125 feet long and consists of a 14-foot-wide by 9-foot-tall closed box structure. The existing culvert has a 53-degree skew from the SW Walker Road centerline and is located approximately 190 feet east of the intersection of SW Walker Road and SW Far Vista Drive. The existing North Johnson Creek culvert on SW Far Vista Drive consists of a 14-foot-wide by 9-foot-tall closed box structure. The existing culvert is located approximately 52 feet south of the centerline of SW Walker Road and is approximately 150 feet long. The existing culverts will be replaced with 20-foot-wide by 10-foot-tall precast box culverts. The SW Walker Road culvert replacement will be 143.8 feet in length and the SW Far Vista Drive culvert replacement will be 70 feet in length.

2.0 PURPOSE AND SCOPE

The scope of our geotechnical services was to explore the subsurface conditions along the bridge alignments to provide geotechnical engineering recommendations for design and construction of the replacement bridges and culvert replacement. The specific scope of services included the following:

Reviewed readily available geologic maps and water well logs that cover the site vicinity.

Reviewed our in-house files for geotechnical information from nearby projects.

Coordinated and managed the field investigation, including scheduling drilling contractors, utility locaters, and GeoDesign staff.

Drilled four borings to depths between 111.5 and 121.5 feet BGS near the proposed abutment locations of the proposed bridge replacements along SW Murray Boulevard and SW Walker Road.

Drilled four borings to depths between 26.5 and 31.5 feet BGS near the existing North Johnson Creek culverts on SW Walker Road and SW Far Vista Drive.

Performed infiltration testing in three hand auger borings at depths between 4.5 and 5 feet BGS.

2 WashCo-56-03:062320

Maintained a detailed log of each exploration, visually classified the soil encountered, collected soil samples as appropriate for the soil conditions encountered, and observed groundwater conditions in each exploration.

Conducted the following laboratory tests on soil samples collected from the explorations:

Seventy-six moisture content determinations in general conformance with ASTM D2216 Four particle-size analyses in general conformance with ASTM D1140 Eight Atterberg limits tests in general conformance with ASTM D4318 Three sieve with hydrometer tests in general conformance with ASTM C117, ASTM C136, and ASTM D422 Provided geotechnical engineering construction recommendations for site preparation, structural fill compaction criteria, wet/dry weather earthwork procedures, and retaining walls.

Provided deep foundation recommendations for the proposed bridges, including bearing capacity, seismic design parameters, and foundation settlement.

Provided recommendations for retaining wall construction, including preliminary global stability and soil parameters.

Provided recommendations for design and construction of precast box culverts, including factored bearing resistance and estimated settlements.

Prepared this report summarizing the results of our subsurface exploration and laboratory testing programs and presenting our geotechnical recommendations and conclusions.

3.0 SITE CONDITIONS

3.1 GEOLOGIC SETTING

The site is located in the Tualatin Basin of the Puget Sound-Willamette Valley physiographic province, a tectonically active lowland located along the convergent Cascadia margin (Orr and Orr, 1999). The Tualatin Basin is formed by a gentle syncline between the uplifted Coast Ranges to the west, the Chehalem Mountains to the south, and the Tualatin Mountains to the north and east. The Tualatin Mountains have been uplifted along northwesterly oriented faults, including the steeply dipping Portland Hills fault located along the eastern flank of the mountains.

The near-surface geologic unit mapped at the site is the fine-grained facies of the Quaternary flood deposits (Madin, 1990). The unit consists of unconsolidated silt and sand deposited by catastrophic floods associated with the sudden release of waters from glacial Lake Missoula during the late Pleistocene Age. Many dozens of these Missoula Floods occurred between approximately 15,500 and 12,500 years ago (and perhaps during earlier glaciations). Flood waters several hundred feet deep swept out of the Columbia Gorge and over the lowlands of the Portland area. The thickness of the flood deposits in the site vicinity is approximately 40 to 50 feet.

Underlying the Quaternary flood deposits are a thick accumulation of moderately to poorly lithified mudstone and sandstone mapped as the Sandy River Mudstone equivalent (Madin, 1990). These were deposited in a large delta where the ancient Columbia and Willamette rivers converged, probably during the middle Miocene to Pleistocene (approximately 6 million to 1 million years ago). The thickness of the unit in the site vicinity is approximately 250 feet.

3 WashCo-56-03:062320

Basement rocks underlying the Sandy River Mudstone equivalent in the site vicinity consist of the Miocene Columbia River Basalts, emplaced approximately 17 million to 6 million years ago in the Portland area (Madin, 1990). The Columbia River Basalts are exposed in the Tualatin Basin in the highlands surrounding the valley and in a group of mountains south of the site, which include Cooper Mountain and Bull Mountain. The Columbia River Basalts consist of thick flows of basalt erupted from fissures in eastern Oregon, Washington, and western Idaho that traveled down the ancient Columbia River Gorge to fill the lowland areas around Portland.

3.2 SURFACE CONDITIONS

The site is located near the intersection of SW Murray Boulevard and SW Walker Road. The project limits extend from the intersection to approximately 1,900 feet east and 720 feet west along SW Walker Road and from the intersection to approximately 800 feet north and 1,250 feet south along SW Murray Boulevard. SW Murray Boulevard will be widened to the east approximately 15 to 20 feet and will require up to 5 to 6 feet of fill, and SW Walker Road will be widened to the north and south approximately 18 feet and will require up to 6 feet of fill. The Nike Campus berm is located south of SW Walker Road east of the intersection and west of SW Murray Boulevard south of the intersection. A retail facility is located at the northwest corner of the intersection and residential apartments are located at the northeast corner. Residential housing is located both north and south of SW Walker Road east of the intersection and east of SW Murray Boulevard south of the intersection. The ground surface in the area of the proposed fill locations is relatively flat beyond the toe of the existing roadway slopes. Figure 2 shows the topography and features at the site.

3.3 SUBSURFACE CONDITIONS

3.3.1 Explorations

The subsurface conditions were explored at each bridge location by drilling a boring near each bridge abutment. Borings B-1 and B-2 were drilled to depths between 111.5 and 121.5 feet BGS for the Murray Boulevard bridge. Borings B-3 and B-4 were drilled to a depth of 121.5 feet BGS for the Walker Road bridge. Borings B-5, B-6, B-7, and B-8 were drilled to depths between

26.5 and 31.5 feet BGS for the culvert extension under SW Walker Road and the culvert replacement under SW Far Vista Drive.

The approximate locations of the explorations are shown on Figure 2. A detailed description of the subsurface exploration program and the exploration logs are presented in Appendix A.

Laboratory testing to determine physical and engineering properties of the soil at the site was completed on select samples from the explorations. Descriptions of the laboratory procedures and test results are presented in Appendix A. The general soil conditions encountered in our explorations are described below.

3.3.2 Subsurface Conditions

3.3.2.1 Murray Boulevard and Walker Road Bridges

The pavement sections near the abutments for the Murray Boulevard and Walker Road bridges consist of approximately 7 to 21 inches of AC underlain by approximately 3 to 12 inches of aggregate base. Beneath the pavement section is approximately 3.5 to 7 feet of gravel, silt, or clay fill with variable sand content and trace organics. Underlying the fill is soft to very stiff alluvial clay and silt. The alluvial silt and clay are generally brown to gray with trace organics and

4 WashCo-56-03:062320 variable sand content. Stratified beds of wet, silty sand less than 1 foot in thickness were present within the silt and clay. The silt and clay are moist to wet and have low plasticity. The recent alluvium extends to depths of 55 to 60 feet BGS in the borings. Laboratory testing of the silt and clay indicates that their moisture content ranged from 26 to 45 percent at the time of explorations. The recent alluvial clay and silt generally have low strength and high compressibility.

Beneath the alluvial silt and clay is Sandy River Mudstone consisting of stiff to very stiff silt and clay. The mudstone is gray to brown in color, generally moist, and ranges from low to high plasticity. The mudstone extends to the maximum depth explored in the borings completed near Murray Boulevard bridge. Medium dense to dense, silty sand between depths of approximately 100 and 115 feet BGS were present at the Walker Road bridge. Laboratory testing indicates that the moisture content of the Sandy River Mudstone ranged from 24 to 45 percent at the time of our explorations.

3.3.2.2 North Johnson Creek Culverts

The pavement sections near the existing North Johnson Creek culvert on SW Far Vista Drive consist of approximately 4.8 to 8.8 inches of AC underlain by approximately 10 to 15 inches of aggregate base. Beneath the pavement section in boring B-5 is approximately 3.5 feet of silt fill.

Beneath the fill or pavement section in borings B-5 and B-6 are alluvial silt and clay consistent with the alluvial soil described for the Murray Boulevard and Walker Road bridges. The alluvial silt and clay extend to the maximum depths explored of 31.5 and 26.5 feet BGS in borings B-5 and B-6, respectively.

The pavement sections near the existing North Johnson Creek culvert on SW Walker Road consist of approximately 10 inches of AC underlain by approximately 11 to 14 inches of aggregate base.

Beneath the pavement section in boring B-7 is approximately 13.5 feet of gravel fill. The gravel is likely associated with backfilling of the existing culvert. Beneath the fill or pavement section in borings B-7 and B-8 are alluvial silt and clay consistent with the alluvial soil described for the Murray Boulevard and Walker Road bridges. The alluvial silt and clay extend to the maximum depths explored of 26.5 and 31.5 feet BGS in borings B-7 and B-8, respectively.

3.3.3 Groundwater

Groundwater could not be directly measured due to the presence of drilling fluid, but increased soil moisture content was observed within 6 to 12 feet of the existing ground surface in most borings. The depth to groundwater will fluctuate in response to seasonal weather and changes in surface topography, and sites within close proximity to the creeks will likely experience groundwater levels consistent with nearby Cedar Mill Creek and North Johnson Creek water levels.

3.4 INFILTRATION TESTING

Infiltration testing was performed in borings HA-1, HA-2, and HA-3. The bases of the borings were saturated prior to testing the infiltration rate using falling-head procedures and a 3-inch-diameter standpipe. Low-head conditions were tested by maintaining the height of the water

5 WashCo-56-03:062320 column at less than 2 feet. Infiltration tests were performed at depths of 5, 4.6, and 4.5 feet in HA-1, HA-2, and HA-3, respectively. Negligible infiltration was observed at all test locations and infiltration systems are not feasible within the project limits.

4.0 LABORATORY TESTING

In addition to the moisture content testing, four particle-size analyses, eight Atterberg limits tests, and three sieve with hydrometer tests were completed on samples collected from the explorations. Descriptions of the laboratory procedures and test results are presented in Appendix A.

5.0 CONCLUSIONS

Based on the results of subsurface explorations, laboratory testing, and engineering analyses, the project is feasible from a geotechnical perspective. The following factors will have an impact on design and construction of the proposed bridge and culvert improvements:

The proposed bridge structures should be supported on driven piles advanced into the stiff to very stiff silt and clay. Pile resistance should be verified with dynamic testing.

The subgrade soil in the areas of the proposed culvert replacements is suitable to support the precast box culvert sections, provided weak and unsuitable subgrade materials are removed and replaced with structural fill and a 2-foot-thick layer of granular material is provided below the culverts.

The near-surface soil at the site consists primarily of fill and native silt and clay. The silty soil can become disturbed during the wet season with repeated heavy traffic. The contractor should protect the subgrade areas from disturbance during construction.

6.0 DESIGN

6.1 PERMANENT SLOPES

Permanent fill slopes on the site should not exceed a gradient of 2H:1V, unless specifically evaluated for stability. Slopes should be planted with appropriate vegetation or armored to provide protection against erosion as soon as possible after grading. Surface water runoff should be collected and directed away from slopes to prevent water from running down the face of the slope.

6.2 FILLS

New approach fills may be required to widen the bridge structures. Based on review of the plans, approximately 5 to 6 feet of fill will be placed on the east side of SW Murray Boulevard to widen the bridge approach and approximately 6 feet of fill will be placed on the south side of SW Walker Road for the bridge approach widening. The fills are confined to the eastern 25 feet of the approach on the Murray Boulevard bridge and the southern 15 feet of the approach on the Walker Road bridge and will not significantly affect the entire approaches. We recommend that hardscapes such as pavement and sidewalks be delayed until after settlement is complete. We estimate that settlement should essentially be complete within three months after fill placement.

6 WashCo-56-03:062320

6.3 PRECAST BOX CULVERT RECOMMENDATIONS

Based on the project information, the SW Far Vista Drive replacement culvert invert elevation will be 170.88 feet, with the channel bottom approximately 4 feet above the invert elevation. The SW Far Vista Drive culvert will be 70 feet in length. We understand that approximately 2 feet of cover will be placed above the top of the SW Far Vista Drive culvert. The invert of the Walker Road culvert replacement will be located northwest of the existing culvert and the outlet will be in the same location as the existing structure. The proposed invert elevation is 171.32 feet, with

3.8 feet of fill above the invert elevation.

Based on our subsurface explorations, the culverts will be founded in medium stiff to stiff silt.

We recommend that a minimum 2-foot-thick layer of crushed rock be placed below the box culverts. To prevent the migration of fines, we recommend that a geotextile separation fabric be placed between the native fine-grained soil and the crushed rock granular fill. The box culverts should be designed using a factored bearing resistance of 3,000 psf with two feet of granular material provided below the culvert. Lateral earth pressures of the box culvert should be calculated using an at-rest earth pressure of 55 pcf for granular fill material placed above the groundwater elevation and 20 pcf for granular fill below the groundwater level. Traffic surcharge loads equivalent to 2 feet of soil or 250 pcf should be applied to the culvert. Surcharge pressures for the fill material above the top of the culvert should be calculated using a unit weight of 125 pcf.

The estimated total primary settlement of up to 2 inches is anticipated for the proposed precast box culvert, retaining walls, and approach embankment fill. Differential settlement is expected to be approximately half of the total. The settlement of the box culvert should essentially be complete within 60 days after fill placement.

We understand that a concrete saddle will be constructed above a 15-inch-diameter waterline that is present under the proposed Walker Road Culvert location. We recommend that the footings for the saddle structure be designed using a factored bearing resistance of 2,000 psf.

We recommend that the fill material placed between the bottom of the culvert and the top of the concrete saddle structure consist of controlled density fill materials.

6.4 SEISMIC CONSIDERATIONS

6.4.1 Geologic Hazards

6.4.1.1 General

In addition to ground shaking, site-specific geologic conditions can influence the potential for earthquake damage. Deep deposits of loose or soft alluvium can amplify ground motions, resulting in increased seismic loads on structures. Other geologic hazards are related to soil failure and permanent ground deformation. Permanent ground deformation could result from liquefaction, lateral spreading, landsliding, and fault surface rupture. The following sections provide discussion regarding potential seismic hazards.

6.4.1.2 Fault Surface Rupture

Fault surface rupture is a seismic hazard that includes abrupt, large, differential ground movements that can cause damage to structures located on a known fault. Based on USGS

7 WashCo-56-03:062320

(2017) mapping, the closest mapped fault is the Oatfield fault, which is located approximately 3 miles northeast of the site. Due to the distance of the closest mapped fault, fault surface rupture is not a design consideration for the project.

6.4.1.3 Liquefaction

Liquefaction can be defined as the sudden loss of shear strength in a soil due to an excessive buildup of pore water pressure. Liquefied soil layers generally follow a path of least resistance to dissipate pore pressures, often resulting in sudden surface settlement, sand boils or ejections, and/or lateral spreading in extreme cases. Clean, loose, uniform or silty, fine-grained, saturated sand is particularly susceptible to liquefaction.

We performed a liquefaction analysis using the Simplified Procedure per recommendations in Section 6.5.2.2 of the ODOT GDM (2018) for sites having potentially liquefiable soil. Based on subsurface conditions, laboratory testing, and analysis, it is our opinion that liquefaction of sand and strain-softening of fine-grained soil will be negligible under the design levels of shaking.

6.4.1.4 Lateral Spreading

Lateral spreading is a liquefaction-related seismic hazard. Areas subject to lateral spreading are typically gently sloping or flat sites underlain by liquefiable sediment adjacent to an open face (such as riverbanks or sloughs). Liquefied soil adjacent to open faces may “flow” in that direction, resulting in surface cracking and lateral displacement toward the open face. The magnitude of lateral spreading decreases with distance from the open face. Because liquefaction is negligible, lateral spreading is not a design consideration.

6.4.2 General

According to the ODOT GDM (2018), new bridges west of U.S. Highway 97 are to be designed for a two-level seismic design criterion. The criteria are presented below:

“Life-Safety” Design Criteria:

Under this level of shaking, the bridge and approach structures, foundation and approach fills must be able to withstand the design forces and displacements without collapse of any portion of the structure and also be consistent with the Life Safety seismic design criteria described below and in the current ODOT BDDM. In general, bridges that are properly designed and detailed for seismic loads can accommodate relatively large deflections without the danger of collapse.

Ground motions for this criterial have an average return period of 1000 years (7% probability of exceedance in 75 years). This level of services is typically what has been provided for bridge design prior to the 2018 of the ODOT GDM.

“Operational” Design Criteria:

This criterion is associated with bridges remaining in service following a level of ground shaking associated with a full-rupture Cascadia Subduction Zone Earthquake (CSZE). For the Operational performance level, bridges and approach fills are designed to provide access for emergency vehicles associated with a full rupture Cascadia Subduction Zone Earthquake (CSZE). Some

8 WashCo-56-03:062320 structural damage is anticipated but the damage should be repairable and the bridge should be able to carry emergency vehicles immediately following the earthquake. This holds true for the approach fills leading up to the bridge.

Approach fill settlement and lateral displacements should be minimal to provide for immediate emergency vehicle access for at least one travel lane. As a general rule of thumb, an estimated lateral embankment displacement of up to 1 foot is considered acceptable in many cases as long as the “operational” performance criteria described above can be met and the structure foundations are adequately designed to withstand the soil loads resulting from the lateral displacements. Vertical settlements on the order of 6” to 12” may be acceptable depending on the roadway geometry, anticipated performance of the bridge end panels and the ability of bridge foundation elements to withstand any imposed downdrag loads.

We note that the Oatfield fault is located approximately 3 miles from the proposed bridge replacements. As described in AASHTO LRFD (2017), a site-specific seismic analysis is required for bridges within 6 miles of an active fault. Based on Sommerville et al. (1997), the spectral acceleration factor associated with the worst-case earthquake scenario is 1.0 between periods of

0.0 seconds and 0.6 second and 1.5 at a period of 5.0 seconds. Based on our experience, the proposed single-span bridges have low seismic requirements and a fundamental period below

0.6 second; therefore, a site-specific seismic analysis is not necessary for the proposed bridge replacements.

6.4.3 Seismic Design Parameters

Due to the presence of liquefiable soil at the site, a site-specific ground motion response may be appropriate. Based on our experience as well as the discussion in Section 6.5.1.4 of the ODOT GDM (2018), the liquefied response spectra are typically lower than non-liquefied response spectra for periods less than approximately 1.0 second. Based on our experience, the proposed single-span bridges will have a fundamental period of less than 1.0 second. Therefore, we did not complete a site-specific ground motion response and conservatively used the code-based spectra parameters.

Earthquake ground motions at the site were evaluated using seismic data from the 2016 ODOT Seismic Design maps. Peak bedrock accelerations for return periods of 500 years and 1,000 years for the “Life Safety” criteria are reported in Table 1. Based on geologic mapping and our explorations, the recommended seismic design Site Class is D. The “Life Safety” seismic design parameters for Site Class D soil are provided in Table 2.

9 WashCo-56-03:062320

Table 1. ODOT Seismic Design Parameters for Bedrock “Life Safety” Criteria

Parameter Value

Site Class (Bedrock) B

Return Period

500 years 1,000 years

PGA 0.178 g 0.278 g

Ss 0.387 g 0.605 g

S1 0.136 g 0.227 g

Table 2. ODOT Seismic Design Parameters for Soil “Life Safety” Criteria

Site Class (Soil) D

500 years 1,000 years

FPGA 1.444 1.322

Fa 1.490 0.316

Fv 2.238 2.416

SDS 0.384 g 0.127 g

SD1 0.211 g 0.325 g

As 0.257 g 0.368 g

Peak bedrock accelerations for the ”Operational” criteria are reported in Table 3. Based on geologic mapping and our explorations, the recommended seismic design Site Class is D. The “Operational” seismic design parameters for Site Class “D” soil are provided in Table 4.

10 WashCo-56-03:062320

Table 3. ODOT Seismic Design Parameters for Bedrock “Operational” Criteria

Site Class (Bedrock) B

CSZE

PGA 0.155 g

Ss 0.292 g

S1 0.162 g

Table 4. ODOT Seismic Design Parameters for Soil “Operational” Criteria

Site Class (Soil) D

CSZE

FPGA 1.490

Fa 1.566

Fv 2.276

SDS 0.305 g

SD1 0.246 g

As 0.231 g

6.5 FOUNDATION ANALYSIS AND DESIGN RECOMMENDATIONS

6.5.1 General

Based on correspondence with WHPacific, Inc., 16-inch-diameter by 0.5-inch-wall thickness, closed-ended pipe piles will be used to support the bridge abutments. The Walker Road bridge will have 15 piles, and the Murray Boulevard bridge will have 18 piles. The Strength 1 and Service 1 AASHTO LRFD (2017) loads for Bents 1 and 2 are shown in Table 5.

11 WashCo-56-03:062320

Table 5. AASHTO LRFD (2017) Service 1 and Strength 1 Loads

Bridge Maximum Service 1 Limit

State Load per Pile (kips)

Maximum Strength 1 Limit State Load per Pile

(kips)

Murray Boulevard 369 527

Walker Road 317 455

1. Loads are based on information from WHPacific.

Based on information provided by the design team, the scour depth at both bridge crossings is estimated to be negligible. The following sections provide recommendations for driven piles in accordance with AASHTO LRFD (2017).

6.5.2 Downdrag

New approach fills may be required to raise the new bridges above design flood elevations. The majority of the fills will be negligible at the bridge abutments, with the exception of the east side of the Murray Boulevard bridge and the south side of the Walker Road bridge, where a maximum of 6 feet of fill will be placed for the widened portion of the bridge approaches. Based on the limited area that will receive new fills, we estimate that settlement will essentially be negligible at the pile foundation locations. Therefore, downdrag loads were not applied in our pile analyses.

6.5.3 Driven Piles

6.5.3.1 Driven Pile Axial Capacities

Sixteen-inch-diameter, closed-ended pipe piles will be used to support the bridges. Analysis was completed to determine factored axial resistances and total nominal resistance during driving for individual piles using methods described in AASHTO LRFD (2017). The analysis is based on the following design and construction assumptions:

The factored pile resistances are based on a resistance factor of 0.65 associated with the use of dynamic pile testing and a calibrated wave equation analysis during construction. This corresponds to an approximate factor of safety of 1.54 for soil resistance on the pile.

Without dynamic pile testing, AASHTO LRFD requires that a significantly lower resistance factor of 0.4 be used (an equivalent factor of safety of approximately 2.5), which will result in longer piles.

The axial resistance factors of the piles should meet the requirements of Section 6.5.4.2 of

AASHTO LRFD.

There are no downdrag effects on the pile.

There are no scour effects on the pile.

The minimum pile spacing, clearance, and embedment into the cap as described in this section are adhered to.

Tables 6 and 7 provide estimated minimum tip elevations, estimated minimum tip lengths, and the required axial resistances per pile assuming PDA testing will be conducted on one pile per abutment for the Murray Boulevard and Walker Road bridges. Pile resistance capacity plots for Bents 1 and 2 for the Murray Boulevard bridge are provided on Figures B-1 and B-2, respectively, 12 WashCo-56-03:062320 in Appendix B. Pile resistance capacity plots for Bents 1 and 2 of the Walker Road bridge are provided on Figures B-3 and B-4, respectively, in Appendix B. The plots present the nominal pile resistances, factored axial resistances for the strength limit state, and the factored uplift resistance for single piles. Analysis verified that the service limit loading will induce settlements less than described in the “Deep Foundation Settlement” section.

Table 6. Pile Capacities and Lengths for 16-Inch-Diameter, Closed-Ended Pipe Piles Murray Boulevard Bridge

Bent

LRFD-

Factored

Load (kips)

Nominal Geotechnical Resistance1

(kips)

Pile Cut-Off

Elevation (feet)

Estimated Tip

Elevation (feet)

Estimated Pile Length Below Bent

Cap (feet)

Factored Uplift

Resistance (kips)

Bent 1 527 811 170.7 64 107 462

Bent 2 527 811 170.7 58 113 406

1. Based on a resistance factor of 0.65.

Table 7. Pile Capacities and Lengths for 16-Inch-Diameter, Closed-Ended Pipe Piles

Walker Road Bridge

Factored

Load (kips)

Nominal Geotechnical Resistance1

(kips)

Pile Cut-Off

Elevation (feet)

Estimated Tip

Elevation (feet)

Estimated Pile Length Below Bent

Cap (feet)

Factored Uplift

Resistance (kips)

Bent 1 455 700 172.7 90 83 279

Bent 2 455 700 172.7 79 94 308

6.5.3.2 Wave Equation Analysis

Based on AASHTO LRFD (2017), wave equation analysis should be completed to determine the driving criteria for piles where dynamic testing is not completed. The wave equation analysis should be performed using the computer program WEAP 1987 (or subsequently newer version) based on information provided by PDA testing. PDA testing will be completed by the agency’s representative (GeoDesign) on a minimum of one pile per pier.

6.5.3.3 Wave Equation Method

A suitable hammer shall be selected for driving piles and for performance of wave equation analyses. The 1987 or newer version of WEAP should be used. The wave equation analyses should be conducted by personnel qualified by training and experience to perform this type of work. The analysis should be completed using a triangular distribution.

13 WashCo-56-03:062320

The WEAP analysis should be concurrent and submitted with the Pile and Driving Equipment Data Form. The Agency will approve or reject the pile driving equipment submittal after a review of the wave equation analysis conducted by the contractor. The pile driving hammers should meet the following requirements based on wave equation analysis:

1. The energy of the submitted hammer shall produce a wave equation-predicted blow count between 3 and 15 blows per 25 millimeters (1 inch) for the ultimate capacities, pile lengths, and other conditions specified.

2. The pile stresses indicated by the wave equation analysis at the ultimate pile bearing capacity shall not be greater than the stress at the point of impending damage to the pile. Tensile and compressive stresses in the pile shall be limited to 90 percent of the pile material’s yield strength for the grade of steel specified at any time during pile installation.

Hammers not meeting these requirements should be rejected. The rejected hammers should be replaced with suitable hammers. Use input values for the wave equation analyses according to Tables 8 and 9.

Table 8. Wave Equation Analyses Parameters – Murray Boulevard Bridge

Factored

Load (kips)

Pile Type Pile

Length1

(feet)

Quake (inches)

Damping (seconds/foot) % Skin

(ITYS)

Rndr

(kips)

Skin Toe Skin Toe

Bent 1 527 PP 16X0.5 107 0.10 0.13 0.2 0.15 87 811

Bent 2 527 PP 16X0.5 113 0.10 0.13 0.2 0.15 76 811

1. The pile length shown does not include the stick-up length above bottom of pile cap.

2. The nominal driving resistance, Rndr.

Table 9. Wave Equation Analyses Parameters – Walker Road Bridge

Factored

Load (kips)

Pile Type Pile

Length1

(feet)

Quake (inches)

Damping (seconds/foot) % Skin

(ITYS)

Rndr

(kips)

Skin Toe Skin Toe

Bent 1 455 PP 16X0.5 83 0.10 0.13 0.2 0.15 61 700

Bent 2 455 PP 16X0.5 94 0.10 0.13 0.2 0.15 68 700

1. The pile length shown does not include the stick-up length above bottom of pile cap.

2. The nominal driving resistance, Rndr.

6.5.3.4 Minimum Pile Spacing, Clearance, and Embedment into Cap The efficiency of pile groups should be evaluated when piles are closely spaced and overlapping stresses from adjacent piles are present or when the pile cap is not in contact with the ground.

Driven pile spacing should not be less than 2.5B (diameter) or 30 inches, center to center, to negate pile group effects to the factored resistances.

14 WashCo-56-03:062320

The distance from the side of any pile to the nearest edge of the pile cap shall not be less than 9 inches. The top of piles should extend at least 12 inches into the pile cap after all damaged material has been removed. If the pile is attached to the cap by embedded bars or strands, the pile shall extend no less than 6 inches into the cap.

Where a reinforced concrete beam is CIP and used as a bent cap supported by piles, the concrete cover on the side of the pile shall not be less than 6 inches, plus an allowance for permissible pile misalignment. Where pile reinforcement is anchored in the cap satisfying Section 5.12.4.1 of AASHTO LRFD (2017), the concrete cover may be less than 6 inches.

6.5.4 Deep Foundation Settlement

Settlement of driven steel pipe piles or drilled piles embedded into dense gravel or sand will be negligible beyond the elastic compression of the pile. For design purposes, a maximum settlement of 1 inch should be estimated for deep foundations designed and constructed in accordance with the recommendations in this report.

6.5.5 Lateral Analysis Parameters

Lateral analysis parameters for analyzing the deep foundations with the LPILE computer software program are presented in Tables C-1 and C-2 for the Murray Boulevard bridge and Tables C-3 and C-4 for the Walker Road bridge (see Appendix C).

In addition to the lateral resistance provided by the foundation elements, lateral forces can be resisted by passive pressure of the pile caps. Provided that the pile caps are at least 2 feet below the ground surface, an available passive earth pressure of 350 psf can be used for loading away from Cedar Mill Creek and where the passive pressures do not exert forces on any planned retaining wall face. We recommend that the available passive earth pressure be reduced to 200 psf where the forces will act on the retaining walls. This additional pressure will need to be accounted for in retaining wall design if they are within the influence zone of retaining walls.

6.5.6 Impact Panels

Impact panels should be constructed on embankment structural fill compacted to at least 95 percent of the maximum dry density, as determined by AASHTO T 99. The total post-construction settlement is expected to be less than 1 inch, and differential settlement between parallel edges of the impact panels should be less than 1 inch given the anticipated loads. This assumes that the full duration of the embankment settlement will occur prior to construction of the impact panels. It is anticipated that the permanent contact pressures imposed on the impact panels will be low enough such that resulting additional settlement will not be significant.

6.6 CIP RETAINING WALLS

6.6.1 General

Based on correspondence with Kittelson & Associates and WHPacific, CIP retaining walls will be required at the bridge abutments. The retaining wall design recommendations presented below are based on the following assumptions: (1) the walls are conventional MSE-type construction,

(2) the walls will be less than 15 feet in height, (3) the backfill is granular, (4) the retaining walls are unrestrained and free to rotate, and (5) the appropriate wall surcharges are included in the design as described in this section.

15 WashCo-56-03:062320

6.6.2 CIP Wall Design Parameters

For unrestrained retaining walls, we recommend an active equivalent fluid pressure of 35 pcf be used for design. If retaining walls are restrained from rotation prior to being backfilled, an equivalent fluid pressure of 55 pcf should be used for design. For embedded building walls, a superimposed seismic lateral force should be calculated based on a dynamic force of 6H2 pounds per lineal foot of wall (where H is the height of the wall in feet). The load should be applied as a distributed load with the centroid located at a distance of 0.6H from the base of the wall. Traffic and retained slope surcharges should be included where appropriate as described in the “Wall Surcharges” section. In addition, the foundation soil for CIP and gravity walls should be firm, granular native or fill soil prepared in conformance with the “Construction” section. Backfill behind the walls should consist of granular wall backfill described in this section.

6.6.3 Wall Surcharges

Where traffic loads are located within a horizontal distance from the top of the wall equal to one-half the wall height, the lateral earth pressure shall be increased by a surcharge load using the following equation and Tables 10 and 11:

p = ksheq where, p = constant horizontal earth pressure due to live load surcharge (psf) k = coefficient of lateral earth pressure (should be taken as 0.28) s = total unit weight of soil (pcf), should be assumed as 125 pcf heq = equivalent height of soil for vehicular load (per Tables 10 and 11

Table 10. Equivalent Height of Soil for Vehicular Loading on Abutments Perpendicular to Traffic

Bent Height (feet) heq

(feet)

5 4

10 3

> 20 2

Table 11. Equivalent Height of Soil for Vehicular Loading on Abutments Parallel to Traffic

Bent Height (feet) heq (feet) Distance from Wall Backface to Edge of Traffic

0 feet 1 Foot or Farther

5 5 2

10 3.5 2

> 20 2 2

16 WashCo-56-03:062320

The design equivalent fluid pressure should also be increased for walls that retain sloping soil.

We recommend the following lateral earth pressures be increased using the following factors (Table 12) when designing walls that retain sloping soil.

Table 12. Lateral Earth Pressure Increase Factors for Sloping Soil Backfill

Slope of Retained Soil

(degrees) Lateral Earth Pressure

Increase Factor

0 1.00

5 1.06

10 1.12

20 1.33

25 1.52

30 2.27

Seismic forces should be modeled based on the pseudo-static approach developed by the Mononobe-Okobe method. As described in the AASHTO LRFD (2017), a seismic coefficient of one-half of the PGA should be used when analyzing internal stability.

6.6.4 Temporary Cuts

Temporary cuts for retaining walls should be completed as described in the “Excavation” section.

6.6.5 Wall Foundations

Retaining wall foundations may be supported on firm native soil or organics-free fill soil provided the foundation subgrade soil is evaluated as described in the “Site Preparation” section. Footings supporting the proposed retaining walls should be designed based on a net allowable bearing pressure of 2,000 psf. The weight of the footing and overlying backfill can be ignored in calculating footing sizes. The recommended allowable bearing pressure applies to the total of dead plus long-term live loads and may be increased by one-third for short-term loads, such as those resulting from wind or seismic forces. Continuous footings constructed on or above slopes should be at least 18 inches wide and at least 24 inches below the lowest adjacent final grade. A minimum horizontal distance of at least 5 feet from the top of the footing shall be provided between the near face of the footing and the nearest face of any finished slope.

Lateral loads on footings can be resisted by passive earth pressure on the sides of the structures and by friction on the base of the footings. Our analysis indicates that the available passive earth pressure for footings is 350 pcf. Adjacent footing elements, pavements, or the upper 12-inch depth of adjacent, unpaved areas should not be considered when calculating passive resistance.

A coefficient of friction equal to 0.40 may be used when calculating resistance to sliding for footings placed on the native or fill gravel.

6.6.6 Wall Backfill and Drains

Granular wall backfill materials placed behind CIP concrete walls should extend at least 1 foot behind the heel of the wall. The granular wall backfill materials should extend to at least 1 foot below the top of the wall where the backfill is level and at least to the top of the wall where the

17 WashCo-56-03:062320 backfill is sloped. Sloping backfill above the retaining wall may consist of general borrow structural fill. Specifications for the fill materials are discussed in detail in the “Materials” section.

A minimum 12-inch-wide by 12-inch-tall zone of drain rock should be placed at the heel of all CIP concrete retaining walls. Perforated collector pipes should be embedded at the base of the drain rock. The drain rock should meet the requirements provided in the “Materials” section. The perforated collector pipes should discharge at an appropriate location away from the base of the wall.

The wall backfill should be compacted as recommended in the “Materials” section. Backfill placed within 3 feet of the wall should be compacted in lifts less than 6 inches thick using hand-operated tamping equipment (such as a jumping jack or vibratory plate compactor).

6.6.7 Construction Considerations

All footing subgrades should be evaluated by the project geotechnical engineer or their representative to confirm suitable bearing conditions. Observations should also confirm that all loose or soft material, organics, unsuitable fill, prior topsoil zones, and softened subgrades (if present) have been removed. Localized deepening of footing excavations may be required to penetrate any deleterious material.

If construction is undertaken during periods of rain, we recommend placing at least 3 to 4 inches of imported granular material over the prepared footing subgrades to help protect the subgrade from disturbance due to the elements and foot traffic. The imported material should meet the specifications for imported granular material as discussed in the “Materials” section.

6.7 PREFABRICATED MODULAR BLOCK WALLS

6.7.1 Wall Geometry

A prefabricated modular block wall is proposed along the south side of the channel realignment of North Johnson Creek. An existing residence is located approximately 22 feet south of the proposed wall location. The wall alignment and section information were provided by Kittelson & Associates. The maximum exposed heights for the wall are expected to range from approximately 4.5 to 6.5 feet with riprap placed at the toe of the wall at a slope of 2H:1V. We understand the riprap is required for scour protection and the riprap layer will have a thickness of 3 feet. The prefabricated modular block wall will require a minimum toe embedment of 2 feet and a wall batter of 1H:8V. The cut for the wall will not be supporting roadways or sidewalks, so a traffic surcharge load should not be applied in design. The contractor should consult the wall manufacturer during construction to assist with block arrangement. Blocks should be arranged so that the entire wall interlocks and acts as a whole and not in independent sections.

6.7.2 Wall Design Parameters

The soil parameters presented in Table 13 should be used for analysis of the prefabricated modular block wall. The material used to backfill behind the walls is discussed in the “Wall Backfill and Drains” section as well as the “Materials” section.

18 WashCo-56-03:062320

Table 13. Wall Design Parameters

Soil Type Unit Weight

(pcf) Friction Angle, ϕ

(degrees) Cohesion

(psf)

Retained Soil – Granular Wall Backfill 1 130 35 0

Retained Soil – Native Silt2 120 28 50

Foundation Soil – Native Silt3 120 28 50

1. Wall backfill for prefabricated modular block, conventional segmental block, and CIP walls should meet the specifications provided in OSSC 00510.12 – Granular Wall Backfill.

2. Retained soil should meet the specifications provided in OSSC 00330.12 – Borrow Material.

3. Foundation soil should be native or fill soil prepared in conformance with the “Site Preparation” section.

6.7.3 Loading Conditions

There are no buildings or roadways located within a horizontal distance from the back of the…