LNAA-CO-26-001 Package A Parking Garage Development Geotechnical Report.pdf

GEOTECHNICAL ENGINEERING REPORT

LVIA Parking Garage, Lehigh Valley International Airport Allentown, Pennsylvania

Schnabel Reference 23150113.000 November 21, 2025

AECOM Technical Services, Inc.

LVIA Parking Garage – Geotechnical Engineering Report

November 21, 2025 Page i Schnabel Engineering, LLC Project 23150113.000 ©2025 All Rights Reserved

GEOTECHNICAL ENGINEERING REPORT

LVIA PARKING GARAGE, LEHIGH VALLEY INTERNATIONAL AIRPORT

ALLENTOWN, PENNSYLVANIA

TABLE OF CONTENTS

1.0 PROJECT DESCRIPTION

1.1 Regional Geology

2.0 EXISTING TERMINAL FOUNDATIONS DOCUMENT REVIEW

3.0 SUBSURFACE EXPLORATION AND LABORATORY TESTING PROGRAM

3.1 Subsurface Exploration Methods

3.1.1 Geophysical Investigation

3.1.2 Test Borings

3.2 Geophysical Results

3.2.1 Frequency Domain Electromagnetic (FDEM)

3.2.2 Seismic Refraction Tomography (SRT)

3.2.3 Multi-Analysis of Surface Waves (MASW)

3.2.4 Soil Laboratory Testing

3.2.5 Index Testing

3.2.6 Compaction Testing

3.2.7 Corrosion Testing

3.2.8 Rock Testing

4.0 SITE GEOLOGY AND SUBSURFACE CONDITIONS

4.1 Site Geology

4.1.1 Variation of Subsurface Conditions Due to Karst

4.2 Generalized Subsurface Stratigraphy

4.2.1 Ground Cover

4.2.2 Stratum A: Fill

4.2.3 Stratum B: Residual Soils

4.2.4 Stratum C: Disintegrated Rock

4.2.5 Stratum D: Epler Formation Bedrock

4.3 Groundwater

4.4 Seismic Site Classification

5.0 KARST HAZARDS AND MITIGATION RECOMMENDATIONS

5.1 Karst Hazards

5.2 Karst Mitigation Recommendations

5.2.1 Site Work

5.2.2 Sinkhole Repair

5.2.3 Limited Mobility Grouting

November 21, 2025 Page ii Schnabel Engineering, LLC Project 23150113.000 ©2025 All Rights Reserved

6.0 GEOTECHNICAL RECOMMENDATIONS

6.1 Site Grading and Earthwork

6.2 Compacted Fill

6.2.1 Fill Settlement

6.2.2 Rock Excavation

6.3 Lateral Earth Pressure Recommendations

6.4 Parking Garage Foundations – Micropiles

6.4.1 Micropile Design

6.5 Spread Footings

6.6 Pedestal Foundations

6.7 Slab-on-Grade

6.7.1 Temporary Slopes

7.0 CONSTRUCTION CONSIDERATIONS

7.1 Site Grading and Earthwork

7.2 Spread Footings and Slab-on-Grade

7.3 Micropile Installation Observation by Foundation Designer (Schnabel)

7.4 Pedestal Foundations

7.5 Rock Excavation

7.6 Engineering Services During Construction

7.7 General Specification Recommendations

8.0 LIMITATIONS

LIST OF TABLES

LIST OF FIGURES

Figure 1: Boring Location Plan Figure 2: Geology Map Figure 3: Karst Features Figure 4: Lateral Earth Pressure Diagram for Design of Below-Grade Walls Figure B-1: Location Plan Figure B-2: FDEM Results Figure B-3: SRT Results Lines 1, 2, and 3 Figure B-4: SRT Results Lines 4, 5, 6, 7, and 8 Figure B-5: MASW Results Lines 1, 2, and 3 Figure B-6: MASW Results Lines 4, 5, 6, 7, and 8

Table 6.1: Estimated Top of Rock Elevations Table 6.2: Recommended Design Parameters for Walls Table 6.3: Micropile Allowable Capacities Table 6.4: Canopy Foundation Loads

November 21, 2025 Page iii Schnabel Engineering, LLC

APPENDICES

Appendix A: As-Built Terminal Plans Appendix B: Geophysical Investigation Results Appendix C: Subsurface Exploration Data Appendix D: Laboratory Results Appendix E: Proposed Micropile Layout (60% Design Development) Appendix F: L-PILE Parameters

November 21, 2025

Mr. Aretin Altmann, PE

1635 Market Street, Suite 1000 Philadelphia, PA 19103

Subject: Geotechnical Engineering Report LVIA Parking Garage, Lehigh Valley International Airport Allentown, Pennsylvania (Schnabel Reference 23150113.000)

Dear Mr. Altmann:

SCHNABEL ENGINEERING, LLC is pleased to submit our geotechnical engineering report for this project. This study was performed in accordance with our executed agreement dated November 6, 2024 (AECOM Project Number 60735239).

Schnabel previously submitted a 30% Geotechnical Design Development Report dated May 30, 2025, and Micropile Plans and Details to AECOM Technical Services (AECOM) as part of the 60% Progress Submission on September 16, 2025.

1.0 PROJECT DESCRIPTION

Lehigh Valley International Airport (LVIA) is a non-hub primary commercial service facility located in Allentown, Lehigh County, Pennsylvania. LVIA is situated approximately 50 miles north of Philadelphia and 75 miles west of New York City. LVIA was originally constructed in 1929 and is currently owned and operated by the Lehigh-Northampton Airport Authority (LNAA). LVIA serves all aviation markets, including general aviation, business, military, and commercial (passenger and cargo).

LVIA has seen a growing volume of passenger-commercial enplanements, and current projections indicate the number will continue to rise. The existing short- and long-term parking lots are at or near capacity during times of heavy air traffic, and LVIA has had to use an overflow lot to accommodate traveler parking. LNAA has initiated a project at LVIA to increase the existing parking capacity at the passenger terminal. We understand LNAA is proceeding with a ±1,250 parking space garage (a net gain of ±975 spaces) alternative.

The proposed parking garage will be located within the existing short-term parking area and will consist of five levels, with the lowest level at grade. The proposed parking garage has a roughly 336 ft by 236 ft footprint and will consist of a pre-cast concrete structure with cast-in-place foundations and shear walls.

The structure will be supported by axially loaded columns and shear walls with axial and lateral loads. An elevator lobby is located at the north end (terminal side) of the garage and will be connected to the

November 21, 2025 Page 2 Schnabel Engineering, LLC terminal via an elevated pedestrian bridge spanning over the existing departure roadway. The proposed garage layout is shown in Figure 1, Boring Location Plan, included at the end of this report.

This report is limited to geotechnical recommendations for the parking garage only; recommendations for the bridge structure are part of a separate design contract.

1.1 Regional Geology

The site is mapped as being underlain by bedrock of the Epler Formation that consists of carbonate-based dolomite rock that is susceptible to solution features resulting in ground subsidence, sinkholes, and/or a highly irregular bedrock surface. Ground subsidence caused by the dissolution of carbonate rock and subsequent loss of soil is referred to as karstic activity, or karst. Karst feature maps from the Pennsylvania Department of Conservation and Natural Resources show surface depressions and sinkholes located within and around the airport. Based on our experience at the airport, the site has a history of sinkholes and surface depressions. A Geologic Map with karst features mapped adjacent to the site is included in Figure 2.

2.0 EXISTING TERMINAL FOUNDATIONS DOCUMENT REVIEW

Schnabel reviewed the existing terminal construction documents dated September 1972 and the As-Built Plan dated April 25, 1974. Modifications to the design were required due to karst conditions encountered during construction, which demonstrates the challenges in working within the site’s karstic geology. The existing terminal building is supported on three foundation types: (1) Type A and C foundations are spread footings on rock with a design bearing pressure of 15 ksf; (2) Type Aꞌ and Cꞌ foundations are spread footings bearing on soil with a 2.5 ksf bearing pressure; and (3) Type B foundations are supported on HP10 piles driven to refusal on bedrock. Modifications noted in the As-Built Plans include:

• Foundations at Columns B2, D8, and D9 were designed to be spread footings, and the As-Built Plans note the foundations consist of both spread footings and driven piles. Some of the foundation dimensions were also modified.

• Foundations at Columns B7, B10, and D6 were designed as driven piles, and the As-Built Plan indicate the foundations were constructed as combined footings and driven piles.

• Two sinkhole repairs are noted on the As-Built Plan, including an 11 cy repair near Column Line 2F and a 2 cy repair between Columns 10A to 10B and H to J.

The As-Built Plan is included in Appendix A. These examples of foundation modifications during construction show the unpredictable nature of working within karst conditions at the site and how important it is to consider the feasibility of different foundation types in these conditions as well as the need to plan and account for modifications needed during construction.

3.0 SUBSURFACE EXPLORATION AND LABORATORY TESTING PROGRAM

Schnabel recommended and performed a two-phase geotechnical investigation. Phase 1 included geophysical investigation methods and a limited number of test borings. The intent of the geophysical investigation is to evaluate the karst conditions and identify potential areas of karst activity across the site, provide a more detailed understanding of the depth to bedrock and variation of rock surface between borings, and map possible areas of near-surface soft soils. The Phase 1 test borings provided soil/rock

November 21, 2025 Page 3 Schnabel Engineering, LLC information at specific points to help initially correlate with the geophysical data. The Phase 2 borings were performed to obtain information to build upon the Phase 1 data to further advance the foundation design, including exploring areas where the geophysical data shows potential karst features and/or varying soil/rock conditions.

3.1 Subsurface Exploration Methods

3.1.1 Geophysical Investigation

Schnabel personnel performed a geophysical investigation at the site consisting of three technologies between December 2 and 12, 2024: frequency domain electromagnetic (FDEM), seismic refraction tomography (SRT), and multi-analysis of surface waves (MASW). FDEM was collected over approximately 4 acres of the site. Data was collected along 8 seismic lines, named Lines 1 through 8, for a total of about 3,000 lf. Details of the geophysical program are described in Appendix B. The FDEM and seismic locations are included on Figure B-1 in Appendix B. Geophysical results were correlated with both the Phase 1 and Phase 2 test boring results and are described below.

The data collection locations and the proposed parking garage layout provided by AECOM are also shown on Figure B-1.

3.1.2 Test Borings

Our subcontractor, Sano Drilling, Inc. of Sewell, New Jersey, drilled a total of 34 test borings, including 17 test borings in Phase 1 between December 2 and 13, 2024, and 17 testing borings in Phase 2 between July 30 and August 29, 2025. The Standard Penetration Test (SPT) was performed at selected depths in the borings. The Borings were named GB-1 to GB-33. Appendix C includes specific observations, remarks, and logs for the borings; classification criteria; drilling methods; and sampling protocols. Figure 1 indicates the approximate test boring locations. We will retain soil samples up to 45 days beyond the issuance of this report, unless you request other disposition.

The SPT samples were obtained using a hydraulically driven automatic-trip hammer (ATH). Most correlations with SPT data are based on N-values collected with a safety hammer. The energy applied to the split-spoon sampler using the ATH is about 33% greater than that applied using the safety hammer, resulting in lower N-values. The hammer blows shown on the boring logs are uncorrected for the higher energy; however, we correct SPT N-values for the higher energy when using N-values in our analyses.

3.2 Geophysical Results

3.2.1 Frequency Domain Electromagnetic (FDEM)

The results from the FDEM survey are shown on Figure B-2. The response is heavily affected by non-geologic features such as underground and above ground utilities, reinforced barriers, and other metallic objects; therefore, interpretations about soil and rock from the FDEM results can be limited in those conditions. We have called out linear features on the results that may correspond to utilities or other cultural features. We observed a low-conductivity zone on the northwest portion of the site that could be indicative of shallow rock. This area correlates well with the seismic and test boring results (detailed below) that interpret shallow rock in this area.

November 21, 2025 Page 4 Schnabel Engineering, LLC

3.2.2 Seismic Refraction Tomography (SRT)

The SRT results are shown as color-coded subsurface profiles of P-wave velocity for Lines 1 to 3 on Figure B-3 and Lines 4–8 on Figure B-4. SRT results are used to interpret the depth to rock. Borehole information is overlaid on the profiles, including soil, rock, areas of low recovery, N-values, recovery, and rock quality designation (RQD) for the rock. Borings that encountered zones of less than 50% of rock core recovery are shown as a hatched diagonal pattern.

Based on our experience in similar geologic settings and the available borehole information, we chose the 2,500 ft/s P-wave contour line to represent the interpreted top of weathered rock and the 4,000 ft/s P-wave contour to represent the interpreted top of less weathered and/or more competent rock. The 2,500 ft/s contour is highlighted as a dotted line, and the 4,000 ft/s contour is highlighted as a square dot line on the profiles on Figures B-3 and B-4.

Line 1 SRT results show a low-velocity zone located between a distance of 60 and 150 ft at about EL 335 ft, with a portion of possible overhanging weathered rock at about a distance of 100 ft; this could indicate a potential karst feature. Boring GB-2 is at this location and shows that the RQD decreases from 100% at the top of rock to 30% at the bottom of the hole where the low-velocity zone is located. This suggests that this feature could be a zone of increased weathering, although the borehole does not rule out a void (open or soil filled) or overhanging rock at the depth identified in Line 1. There is also a zone of lower P-wave velocity at EL 310 to 320 ft along Line 1 between about 200 and 275 ft distance that may be related to increased fracturing or weathering within the rock. Boring GB-3 encountered low recovery between about EL 320 and 330 ft that supports this trend.

Line 2 SRT results show a variable depth to weathered and less weathered rock. At the southern end of Line 2, between a distance of 475 and 500 ft, the depth to interpreted weathered rock decreases from about EL 350 to 320 ft. Line 3 SRT results show a deeper depth to weathered rock and lower P-wave velocity within the weathered rock from about a distance of 0 to 50 ft. This is not reflected in the RQD of Boring GB-21, which is 20 ft offset from Line 2 and may be caused by an adjacent lower-quality rock zone.

Figure B-4 includes SRT Lines 4 to 8. The SRT results for Lines 4, 5, 6, and 7 show variable depth to weathered and less weathered rock, ranging from about 7 to about 45 ft deep.

The results from SRT Line 4 and GB-16 show a discrepancy; rock was encountered in GB-16 at about EL 310 ft, with low blow counts (N value) above rock. This is about 10 ft deeper than would be expected based on the SRT Line 4 results. We interpret this discrepancy to be due to the 17-ft offset of the borehole from the seismic line and the expected variable nature of the karst rock.

SRT Line 8 shows highly variable conditions from about a distance of 150 ft to the end of the line (300 ft), with interpreted top of rock ranging from about a 5-ft depth to more than 40 ft deep. Borings GB-28 and GB-30 results correlate well with the LSRT Line 8 results, indicating more variable and poorer quality rock based on recovery and RQD. On Line 8 at about a distance of 240 ft, the results show a low-velocity zone at about EL 330 ft that we interpret as increased weathering or fracturing or possible soil or void within the rock.

November 21, 2025 Page 5 Schnabel Engineering, LLC

We observe a trend with the shallowest weathered bedrock in the northwest corner of the site, as was suggested by the FDEM results. We also observe deeper weathered rock to the south of the area with shallow bedrock (near GB-2 and GB-4). Other areas with the deepest weathered bedrock are located in the south center portion of the site, as indicated on Line 2; in the northeast corner along Line 3; and in the northwest area where Lines 1 and 5 intersect where deeper rock was observed in GB-4 and GB-8.

3.2.3 Multi-Analysis of Surface Waves (MASW)

The MASW results are shown as color-coded subsurface profiles of S-wave velocity for Lines 1 to 3 on Figure B-6 and for Lines 4 to 8 on Figure B-7. While SRT results are used mainly to estimate the depth to rock and typically have a higher resolution, MASW results are used to estimate areas with soft soils and softer or stiffer rock below the top of rock. MASW results also tend to have a lower resolution and represent more of a bulk shear-wave velocity in the measured areas rather than a precise velocity at a specific location. SRT results are shown in P-wave velocities, while MASW results are shown in shear wave (S-wave) velocities.

For reference, the 2,500 and 4,000 ft/s P-wave contours from the SRT results (from Figures B-3 and B-4) are shown overlain on the MASW results in Figures B-6 and B-7. The zones with the highest shear-wave velocity from the MASW results correspond well with the zones with the highest P-wave velocity from the SRT results, with these areas generally corresponding to the interpreted less weathered (more competent) rock.

The S-wave velocity of 800 ft/s is highlighted on the MASW results as a grey hatched line, and the zones with S-wave velocities less than 800 ft/s represent softer soils. Typically, based on our experience, very soft and soft soils are indicated by S-wave velocities below about 400 to 500 ft/s. The variability in the MASW results indicates isolated, relatively thin zones of lower S-wave velocity materials, which correlates with soils with lower N-values encountered in the borings. The softer zones vary across the site and range from the near surface to about 25 ft deep.

3.2.4 Soil Laboratory Testing

Schnabel sent samples collected during the subsurface exploration to AECOM Technical Services Inc. in Conshohocken, Pennsylvania, to perform testing. The testing aided in the classification of materials encountered in the subsurface exploration and provided data for use in the development of recommendations for design of foundations and earthwork. The results of the laboratory tests are included in Appendix D and are summarized (for each stratum) in the Site Geology and Subsurface Conditions section of this report. Selected test results are also shown on the boring logs in Appendix C.

3.2.5 Index Testing

We performed index testing on samples collected as part of the exploration to provide soil classifications and to provide parameters for use with published correlations with soil properties. Index testing included performing natural moisture content, Atterberg Limit, gradation tests, and hydrometer analysis on 31 jar and 4 bulk sampls of soil representing Strata A and B.

November 21, 2025 Page 6 Schnabel Engineering, LLC

3.2.6 Compaction Testing

We performed Standard Proctor compaction testing to evaluate compaction characteristics and to provide soil parameters for pavement design. Testing was performed on two samples representing Stratum A, one sample representing Stratum B, and one sample combined from Strata A and B.

3.2.7 Corrosion Testing

We performed tests for pH, sulfides, redox potential, and resistivity testing on two samples representing Strata A and B.

3.2.8 Rock Testing

We performed compression point load testing of 32 rock specimens representing Stratum D to provide information regarding the strength of the rock encountered on site.

4.0 SITE GEOLOGY AND SUBSURFACE CONDITIONS

4.1 Site Geology

A review of geologic information from the Pennsylvania Department of Conservation and Natural Resources indicates the project site is underlain by the Epler Formation of the Ordovician Period. The Epler Formation is described as a thick bedded, finely to very finely crystalline, medium gray limestone with yellow dolomitic laminae and interbedded medium, dark gray, finely crystalline dolomite that weathers yellowish gray. The Epler Formation can contain conglomerate, fossil fragments, and oolitic lenses. The thickness of the Epler Formation is noted as up to 800 ft (Markewicz and Dalton 1977). This formation generally has moderately abundant, open and steeply dipping, well to poorly developed fractures; is moderately resistant to weathering; can have a soil overburden of 80 ft or more; and typically has difficult excavation properties due to pinnacles (Geyer and Wilshusen 1982).

4.1.1 Variation of Subsurface Conditions Due to Karst

The limestone and dolomite of the Epler Formation are carbonate bedrock. Carbonate bedrock is subject to dissolution. The dissolution typically results in highly variable subsurface conditions (both vertically and laterally) and commonly includes bedrock fractures adjacent to very hard rock, deep bedrock zones filled with soft soil adjacent to more shallow bedrock, overhanging ledge rock, pinnacled bedrock (i.e., the top of the intact rock surface can vary significantly over short distances), boulders, very soft soils weathered from the bedrock at the soil-bedrock interface, and solution features in the rock such as voids filled with soil, water, or air. These features are described as “karst” features.

Karst features, such as bedrock pinnacles, soft or loose material beneath firmer or stiffer soils, and deep, highly fractured, and weathered zones of bedrock were observed across the site. A Karst Features and Reference Map is included in Figure 3.

4.2 Generalized Subsurface Stratigraphy

We encountered the following stratigraphy during our exploration. We characterized the generalized subsurface stratigraphy based on the exploration and laboratory test data included in the appendices.

November 21, 2025 Page 7 Schnabel Engineering, LLC

4.2.1 Ground Cover

The borings encountered a pavement section ground cover consisting of asphalt and subbase. The asphalt section ranged from 4 to 10 inches thick and was underlain by 3 to 10 inches of subbase. The variation in asphalt and subbase thickness may reflect the history of utility work and modifications to the pavement section. Stratum A: Fill

Fill soils were encountered in in eight borings below the ground cover (GB-8, GB-11, GB-16, GB-19, GB- 20, GB-24, GB-28, and GB-33). The fill was encountered at depths of 0.8 to 12 ft below ground surface (EL 360.2 to 349). The fill was generally described as an orangish or grayish brown lean clay (CL) with varying amounts of sand and gravel. Brown poorly graded sand with silt (SP-SM) was observed within the fill of Boring GB-20. Quartz, asphalt, and calcite fragments were also observed within the fill soil.

Uncorrected SPT N-values within the fill ranged from 5 to 20. One bulk sample of the soil from Stratum A and one bulk sample of Strata A and B soils combined were tested for moisture content, gradation, Atterberg Limits, and hydrometer analysis. The reported natural moisture contents were 14.2 and 14.6%;

Liquid Limits ranged from 31 to 34 for both; Plastic Limits were both 18; and 63.4 and 77.0% passed the No. 200 sieve. Hydrometer analysis reported 29.3 and 32.1% clay content.

4.2.3 Stratum B: Residual Soils

Residual soils are derived through the in-place physical and chemical weathering of the underlying rock.

The residual soils were encountered at depths of 0 to 60 ft (EL 363 to 300). The residual soils were generally described as an orangish or grayish brown, reddish tan, or tan lean clay (CL) with varying amounts of sand and gravel, poorly graded sand with silt or clay (SP-SM/SP-SC), sandy silt (ML), clayey or silty sand (SC/SM), and clayey gravel (GC). Generally, uncorrected N-values within the residual soils ranged from 1 to >50. Thirty-three SPT samples and two bulk samples of the soil from Stratum B were tested for moisture content, gradation, Atterberg Limits; hydrometer analysis was performed on nine samples. Two SPT samples were tested for moisture content only. The reported natural moisture contents ranged from 13.3 to 31.0%; Liquid Limits ranged from 25 to 43; Plastic Limits ranged from 18 to 28; and 24.0 to 93.1% passed the No. 200 sieve. Four samples were found to be non-plastic. Hydrometer analysis reported 20.7 to 41.7% clay content.

Moisture-Density testing (Standard Proctor) was performed on two bulk samples representing this stratum. The samples consisted of sandy lean clay or lean clay with sand and classified as CL in accordance with the USCS classification system. The compaction tests resulted in corrected maximum dry densities of 116.0 and 112.9 pcf at corrected optimum moisture contents of 14.6 and 16.1%. Natural moisture content values of Stratum B soils within roughly 10 ft of existing grade varied from roughly 1 to 13% wet of optimum.

Testing performed to evaluate the corrosion potential of two Stratum B soil samples reported pH values of

7.9 and 7.8, resistivity values of 1,090 and 980 ohm-cm, chloride concentrations of 9.5 and 77.8 ppm, and sulfate concentrations of 82.4 and 450.2 ppm. Based on these test results, the soils of Stratum B exhibit potentially corrosive conditions according to the criteria in PennDOT Publication 15M, DM-4 (2019).

4.2.4 Stratum C: Disintegrated Rock

Disintegrated rock is defined as residual material with SPT N-values between 60 blows per foot and refusal. Refusal is defined as an N-value of 50 blows for a penetration of 1 inch or less. Disintegrated rock

November 21, 2025 Page 8 Schnabel Engineering, LLC was encountered in eight borings: GB-3, GB-6, GB-7, GB-16, GB-23, GB-25, GB-26, and GB-30 at depths ranging from 4 to 45.5 ft below grade (EL 357 to 315.5).

4.2.5 Stratum D: Epler Formation Bedrock

The dolomite bedrock of the Epler Formation was encountered in the borings. The rock cores were generally described as dark gray to gray limestone with streaks of white, weak to strong, severely to slightly weathered, and intensely to slightly fractured. Water return was generally not observed during rock coring, further indicating the presence of fractures, weathered, and/or voided rock conditions.

Laboratory testing of rock samples of Stratum D indicates unconfined compression strength (UCS) values ranging from 1,511 to 28,284 psi with an average of 12,998 psi. Evidence of karst bedrock conditions was observed as demonstrated by the following conditions encountered within the rock:

• Test Borings GB-4/GB-9/GB-12/GB-15/GB-17/GB-19/GB-22/GB-23/GB-29: One or more soil seams ranging in thickness between 0.2 to 6.5 ft were encountered below top of rock, with the soil seams observed at depths ranging between 12 and 62.7 ft below grade (EL 349 and 300).

• Test Boring GB-6 and GB-6A: Core barrel deflection at 10 ft (EL 354) in GB-6 due to possible rock pinnacle. Tool drop in GB-6A through void from 11 to 14 ft below grade (EL 353 to 350).

• Test Boring GB-7 and GB-14: Split spoon and core barrel deflection at 4 to 5 ft in GB-7 (EL 358 to

357) and 23 ft in GB-14 (EL 337) due to possible pinnacle.

• Test Boring GB-8: Encountered several soil seams, including a 15 ft soil seam from 45 to 60 ft (EL 316 to 301); and 1 ft, 5 ft, and 2 ft soil seams at depths of 62 to 63 (EL 299 to 298); 65 to 70 ft (EL 296 to 291); and 76 to 78 ft (EL 285 to 283), respectively. Split-spoon deflection was observed in GB-8 at a 30 ft depth due to a suspected rock pinnacle.

• Test Boring GB-13/GB-28: Tool drop in GB-13 through void from 27.2 to 27.5 ft below grade (EL 331.8 to 331.5), as well as a 1.5 ft soil seam at a depth of 32 to 33.5 ft (EL 327 to 325.5).

Tool drop in GB-28 through possible void from 55 to 57 ft below grade (EL 304 to 302).

• Test Boring GB-20: Encountered a 12 ft soil seam (EL 326.5 to 314.5) below top of rock while drilling.

4.3 Groundwater

We monitored groundwater during drilling and obtained water readings during drilling and upon completion prior to backfilling the test borings. The borings were observed to be dry and groundwater was not observed during the exploration. We did not obtain long-term water level readings because the test borings were backfilled upon completion for safety.

The final design should anticipate the fluctuation of the hydrostatic water table depending on variations in precipitation, surface runoff, evaporation, leaking utilities, and similar factors.

4.4 Seismic Site Classification

We evaluated the Seismic Site Class and Seismic Site Coefficients for this project according to ASCE 7-

16. Our analysis indicates Site Class D based on corrected SPT N-values. The proposed structures are based on Risk Categories III and IV. We recommend the following mapped acceleration parameters:

November 21, 2025 Page 9 Schnabel Engineering, LLC

• Short period parameter (Ss) of 0.169

• 1-second period parameter (S1) of 0.048

• Site amplification factor (Fa) of 1.6

• Site amplification factor (Fv) of 2.4

• Numeric seismic design value at 0.2 seconds (SDS) of 0.18

• Numeric seismic design value at 1.0 second (SD1) of 0.076

5.0 KARST HAZARDS AND MITIGATION RECOMMENDATIONS

5.1 Karst Hazards

As discussed in Section 4.1, the site is underlain by carbonate bedrock that is susceptible to dissolution.

Karst activity resulting from the process of dissolution is a natural process that can be modified and exacerbated by the construction of new features. The proposed development may exacerbate this risk of karst activity both short and long term.

By undertaking the proposed parking garage, Schnabel assumes the Owner’s understanding and acceptance of the risk of exposure to the potential for subsidence due to karst activity during the lifecycle of the proposed facility. We anticipate three primary risks associated with the proposed parking garage project that are related to karst activity:

• Encountering boulders and bedrock pinnacles during construction

• The development of sinkholes and subsidence during construction

• The development of sinkholes and subsidence after construction, and its impact to completed work

Closed depressions and sinkholes are the two main types of cover (ground) subsidence that are visible at the surface in karst areas. Both features, discussed below, are closed drainage systems, meaning that water flows into them and into the underlying soil and rock, which can exacerbate the subsidence.

• Closed Depressions: Closed depressions typically form where granular materials spall or settle into a void gradually, like sand in an hourglass moving downward. Closed depressions generally develop at the ground surface slowly, and there is not a collapse, soil scarp, or visible throat.

• Sinkholes: Sinkholes generally develop in the subsurface over some time period, where material spalls into a void in the underlying rock, sometimes called a “throat,” and a void propagates upward in the soil. Eventually the void reaches a point where the cover soil roof collapses, resulting in a hole at the ground surface with fresh soil exposed at the sidewall or scarp. If rock is shallow, the throat may be visible within the fresh scarp. Over time, if a sinkhole is not remediated, the edges of the feature may soften and collapse inward and be indistinguishable from a closed depression.

The risk of sinkhole development can be elevated during construction when surface drainage is abnormal and changes in stresses on the soil occur from cutting/filling and from the load of the traffic of construction equipment. Following construction, the risk of sinkholes can also be elevated due to changes in the soil cover (i.e., the reduction in the thickness of the soil cover above bedrock can result in more rapid

November 21, 2025 Page 10 Schnabel Engineering, LLC subsurface erosion and infiltration) and in the drainage regime (i.e., concentrated water flow and infiltration of surface water runoff from stormwater facilities [pipes, basins, etc.]).

The formation of sinkholes during and following construction should be anticipated. Recommendations to reduce the risk of and remediate against the effects of karst activity are provided below.

5.2 Karst Mitigation Recommendations

The depth, size, shape, and nature of the subsidence caused by karst activity can vary considerably. In some cases the bedrock surface is exposed, and in other cases only soil is visible. Thus, each subsidence feature should be repaired in accordance with recommendations from the Geotechnical Engineer. As such, the Geotechnical Engineer should be notified if subsidence develops, as each remediation event will depend on multiple factors, including depth to bedrock, size of the subsidence feature, and location with respect to critical structures. Recommendations for repairs are provided below.

5.2.1 Site Work

The following practices are recommended to reduce the potential for the development of sinkholes during the site work construction.

• Site excavations should be performed to minimize the potential for ponding water.

• Earthwork should be sequenced to minimize the time excavations are left open.

• Accumulated water in depressions should be pumped out to avoid any ponding of water.

5.2.2 Sinkhole Repair

When sinkholes or soft zones are observed during construction, soft or collapsed soils should be excavated in an attempt to reveal the throat of the sinkhole. Water may be used to wash pinnacles, open throats, and evaluate drainage. The throat should be plugged as recommended by the Geotechnical Engineer. A description of the sinkhole repair details and logic is as follows:

• Sinkhole Repair or Soft Zone in Soil: In areas where a defined sinkhole throat in rock is not observed, loose materials should be removed and a woven geotextile placed to contain fill materials used to repair the area. Crushed stone may be used to stabilize the base of these areas prior to placing capping soils. Place geotextile over stone, then place low permeability soil defined as a minimum 65% passing the No. 200 sieve, a minimum Liquid Limit of 25, and a minimum Plasticity Index of 10.

• Sinkhole Repair in Rock: Where competent rock sides are encountered and a clear throat of the sinkhole can be identified, the throat should be filled with crushed rock fill followed by the placement of a low-strength concrete seal to limit future infiltration. The required cover materials (asphalt, concrete, soil cover with topsoil/seed, etc.) can be placed once the concrete cap cures.

5.2.3 Limited Mobility Grouting

Limited Mobility Grouting (LMG) is a potential karst remediation and recommended in areas where there are deep and wide zones of unsuitable/unstable material and can be used to:

• Fill voids/displace soft soils

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• Densify soils to improve the bearing capacity

• Plug sinkhole throats at the bedrock surface to prevent downward migration of soil into voids

The LMG technique uses a low mobility (low slump) cement-based grout to densify soil through controlled injection and radial expansion. As part of the grouting program, probe holes are usually extended several feet into rock (or some pre-determined minimum depth) to delineate soft zones and/or voids within the rock mass. Typically, the grouting is performed in a grid pattern of primary holes, followed by secondary (and possibly tertiary) holes at a closer spacing where warranted by the volume of grout takes observed in the primary holes. To limit grout takes, perimeter holes are often completed first to act as a barrier before interior holes are grouted to achieve closure.

We recommend the contract include provisions for this remediation method in the event it is deemed necessary by the Geotechnical Engineer. If deemed necessary, a location-specific grouting program and layout would be developed by the Geotechnical Engineer based on karst feature sizes and depths identified in the field.

6.0 GEOTECHNICAL RECOMMENDATIONS

We based our geotechnical engineering analysis on the information developed from our subsurface exploration and soil laboratory testing, along with the project development plans, site plans, and structural loading furnished to our office. The following sections of the report provide our detailed recommendations.

6.1 Site Grading and Earthwork

The proposed parking garage is expected to require minor (<3 ft) grade change(s) to the site. Fill placement on the project is limited to roughly 6 to 8 ft at the garage ramp, between Column Lines B and C from Column Lines 3.5 to 7. Grade changes (cuts) of roughly 6 inches up to 2.5 ft will generally be required within the garage footprint for slab-on-grade construction. The proposed grading indicates that the site grades are similar to existing grades. Recommendations for compacted fill, subgrade preparation, fill soil requirements, placement and compaction criteria, monitoring fill settlement, and evaluating rock excavation are presented in subsequent sections.

6.2 Compacted Fill

Compacted structural fill should consist of non-organic on-site soils classifying as CL, ML, SC, SM, SW, GC, and/or GM. If off-site borrow materials are needed, these soils should classify as SC, SM, SW, GC, or GM according to ASTM D2487. Fill materials should not contain particles larger than 3 inches. In addition, off-site fill materials should exhibit Liquid Limit and Plasticity Index values of less than 45 and 20, respectively.

Compacted structural fill should be placed in maximum 8 inch thick horizontal loose lifts. Fill should be compacted to at least 95% of the maximum dry density per ASTM D698 (Standard Proctor).

Backfill in excavations, trenches, and other areas that large compaction equipment cannot access should be placed in maximum 6 inch thick lifts. Backfill should meet the material, placement, and compaction requirements outlined above.

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Successful reuse of the excavated, on-site soils as compacted structural fill will depend on their natural moisture contents during excavation. Laboratory test results indicate soils in areas that will be excavated during garage construction are <1 to 13% wet of the optimum moisture content, with the average in situ moisture contents in the upper 10 ft being roughly 3.9 to 6.2 percent wet of the optimum moisture content.

Scarifying and drying of these soils should be anticipated to achieve the recommended compaction.

Drying of these soils will likely result in some delays; could be difficult without sufficient open areas; and may not be possible during cooler, wetter weather. We recommend that the earthwork be performed during the warmer, drier times of the year.

6.2.1 Fill Settlement

The subgrade soils are expected to settle up to ½ inch under the weight of the 6 to 8 ft of proposed ramp fill. Construction of slab-on-grade foundations for the ramp should not begin until settlement has essentially ceased. We recommend that two settlement plates be installed in the ramp fill area to monitor settlement. Recommended settlement plate locations are indicated in Figure 5. Settlement plates should be installed on the fill subgrades prior to placing the compacted structural fill.

Settlement plates should consist of a 24 x 24 x ½ inch steel plate with a 2-inch steel riser pipe welded to the plate’s center. The plate should be placed on the subgrade, and the elevation of the plate and top of the riser recorded before fill placement. As fill operations progress, the Contractor should extend the riser pipe to remain above the fill surface. The elevation of the top of the riser should be recorded immediately before and after attaching an extension. The elevation of the riser should be recorded daily. After completion of the fill, the elevation of the top of the riser should be recorded three times a week.

Settlement readings should be provided to the Geotechnical Engineer for evaluation of when above-grade construction may proceed.

Settlement plate readings should be taken to the nearest 0.005 ft and referenced to a benchmark well beyond the influence of the fill placed. Precautions should be taken to prevent damaging the settlement plates during fill operations. The general Contractor should furnish all labor and materials and perform all operations needed for installation and maintenance of the settlement plates.

6.2.2 Rock Excavation

The test boring data, when compared to proposed foundation grades, indicate that rock may be encountered during pile cap and grade beam excavation. Conditions encountered during excavation may be different from those observed in the test boring data; therefore, we recommend the contract documents include unit prices for rock excavation. The boring data indicate rock excavation methods may be needed. Table 1 provides elevations where shallow top of rock was observed and excavation methods, such as hoe-ramming, could be required. Blasting is not recommended.

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Table 6.1: Estimated Top of Rock Elevations

Boring Number Top of Rock Elevation (ft)

GB-2 352.0

GB-6A 355.0

GB-18 349.0

GB-261 356.5

Notes: GB-26 was completed outside the proposed garage footprint.

Variations in rock conditions should be expected because the rock surface can fluctuate across the site due to karst. Also, the extent of rock excavation will depend on the Contractor’s methods, rock jointing, and rock foliation. The elevations noted above apply for mass excavation of rock on the site. Rock excavation methods may be needed at higher elevations in footing and utility excavations.

A sample definition of rock for excavation specifications is provided below:

For mass excavation, rock is defined as any material that cannot be dislodged by a Caterpillar Model No. D-8 heavy-duty tractor, or equivalent, equipped with a hydraulically operated, single-tooth power ripper without the use of hoe-ramming or blasting. For trench, footing and pit excavations, rock excavation shall be defined in terms of a Caterpillar Model No. 330 hydraulic excavator, or equivalent. This classification does not include material such as loose rock, concrete, cemented gravel, or other materials that can be removed by means other than hoe-ramming or blasting, but which for reasons of economy in excavating, the Contractor chooses to remove by hoe-ramming or blasting. Rock does not include boulders less than one cubic yard in volume. Boulders larger than one cubic yard in volume will be considered rock for payment purposes.

Where the rock cannot be removed with conventional excavation equipment, special means of excavation may be needed. Removal of rock may require the use of blasting, air-powered tools, rock splitters, or large hoe rams.

6.3 Lateral Earth Pressure Recommendations

The proposed parking garage contains a ramp to allow vehicle traffic to move from the ground floor to the second level. The cast-in-place ramp walls will be braced by the structure, preventing movement. The wall-backfilling will take place before the ramp slab(s) are poured; therefore, the Contractor will need to incorporate measures to prevent rotation/deflection of the wall. The ramp walls should be designed considering equivalent fluid pressure, gef, as shown in Figure 4. Equivalent fluid pressure factors are presented in Table 2. Where applicable, the design should consider surcharge loads using a rectangular earth pressure distribution as shown in Figure 4. The surcharge pressure ordinate should be obtained by multiplying the surface surcharge pressure, q, by the surcharge pressure factor in Table 2.

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Table 6.2: Recommended Design Parameters for Walls

Backfill Materials

Equivalent At-Rest Fluid Pressure Factor

(gef)

Surcharge Pressure

Factor On site 55H 0.5

Materials used for backfill for the ramp walls of the proposed parking garage should meet the requirements for compacted structural fill/backfill. The Contractor should exercise care when backfilling adjacent to the ramp wall(s) to not damage the structural integrity of the concrete and the damp-proofing.

Only light, hand-operated equipment should be used to compact backfill against walls. The Structural Engineer of Record should approve the size of the compaction equipment.

6.4 Parking Garage Foundations – Micropiles

Shallow foundations rely on the strength and compressibility properties of the underlying soil for support of a structure. Karst geologic formations inherently have an increased risk (compared to non-karst geology) for subsidence or settlement of shallow foundations due to the nature of carbonate rock. The site’s history of sinkholes, Terminal Building foundation modifications during construction, and conditions encountered in the test borings confirm the unpredictability and high-risk nature of the site’s geology that is not conducive to shallow foundations. Furthermore, the proposed parking garage column loads vary from roughly 780 to 1,820 kips and vertical components of the wall loads vary from 42 to 138 kips/ft.

These large loads would likely result in high bearing pressures for spread footings, with total and differential settlements exceeding tolerable limits. For these reasons, we do not recommend shallow foundations.

Deep foundations may consist of driven piles or various drilled foundation types. Driven piling is problematic in karst due to the highly variable bedrock surface that can result in long pile lengths and piles out of plumb or position tolerance from the pile’s inability to be seated into sloping bedrock. Short pile lengths can result in very low uplift capacities. Driven piles can also refuse on boulders or pinnacled rock ledges in karst that can result in a reduced capacity of the bearing layer should these features yield.

Driven piling also involves noise and vibration during construction.

Drilled foundations may be either large or small in diameter. Large diameter drilled foundations (e.g., drilled shafts) have disadvantages over small diameter drilled foundations (e.g., micropiles) at this site because they require large installation equipment, are less flexible with element placement, and can be problematic during installation in karst geology. Large drilled elements can have issues “biting” into sloping bedrock due to their diameter. Large elements are also less conducive to flexibility in element placement and with the various constraints at this site can be restrictive; therefore, we recommend a small diameter drilled foundation (micropiles) for this project.

Micropiles are a drilled, deep foundation consisting of a steel casing, cement grout, and reinforcing bar(s).

Rock-socketed micropiles are recommended for this site, where the casing extends through the overburden soil to the rock socket. The rock socket consists of the reinforcing bar and cement grout that bond to the rock giving the micropile axial capacity in compression and tension. Micropiles that are designed to be socketed into rock and are installed properly can achieve low settlement tolerances and greatly reduce the risk of settlement of the structure due to karst activity.

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An advantage of using micropiles in karst conditions is that the pile can be advanced through voids, soil-filled voids, and/or broken rock conditions to a competent rock layer. Foundations drilled into rock can be monitored during installation to ensure they are socketed into competent bedrock, as opposed to a driven pile foundation where competent bearing rock beneath the tip elevation cannot be verified. As a small diameter drilled foundation, micropiles are capable of “biting” into sloping bedrock more easily than a large diameter drilled foundation. Furthermore, the small diameter foundation equipment is advantageous when working around the tight site constraints (utilities, adjacent parking areas, etc.) because the micropile equipment has more flexibility and requires less lay-down areas for equipment and material staging compared to larger foundation equipment. Micropile installation equipment can also be used to perform Limited Mobility Grouting (LMG) operations to remediate sinkholes or fill encountered voids. An LMG sinkhole remediation was performed during micropile installation for the LVIA Terminal-Connector project, with minimal downtime in transitioning between micropile installation and LMG operations.

6.4.1 Micropile Design

THA Consulting, Inc. provided Schnabel via email with unfactored design loads at column/wall foundations for the proposed parking garage. Dead loads, live loads, and snow loads were provided for the stair towers, columns, light walls, and base shear loads for both types of walls. Load combinations were evaluated using Section 1605.2 of the 2021 International Building Code (IBC 2021) using allowable stress design. THA Consulting, Inc. also provided tolerable total and differential settlements of 1 and 0.5 inch, respectively.

Calculations were performed to estimate the sizes and lengths of the various micropile elements using loading provided by THA from the 60% submission. We understand the loads will be refined as the design process continues. Calculations were performed using the Micropile Design and Construction Reference Manual (FHWA NHI-05-039) and IBC 2021. Table 3 shows conceptual micropile allowable stress design capacities. The capacities are based on an ultimate grout-to-rock bond stress of 150 psi. This bond stress should be confirmed during construction as part of a load testing program. Details of the load testing program will be outlined in the micropile drawings and specifications. Due to the high structural loads from the garage building, using larger diameter micropiles was considered because larger diameter micropiles have higher capacities, resulting in fewer piles during construction and overall lower construction costs and risks.

Table 6.3: Micropile Allowable Capacities

Micropile Diameter (inches)

Bond Length in Rock

(ft) Allowable Capacity

(kips)

10.75 10-15 275–400 Note: Capacities are for conceptual design purposes only. Details such as casing thickness, reinforcement size, etc. are not specified herein. Refer to design submissions for more details.

A 60% Design Development Micropile Layout, based on 10.75-inch diameter micropiles with 380 kip allowable capacities, is included in Appendix E.

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