B08_Attachment_8-Geotechnical_Investigation_Report.pdf

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REPORT ON

DESIGN LEVEL GEOTECHNICAL INVESTIGATION

STINSON BEACH LIFEGUARD TOWER AND VISITOR SNACK BAR

REPLACEMENT

MARIN COUNTY, CALIFORNIA

by Haley & Aldrich, Inc.

Walnut Creek, California for DHM Design Denver, Colorado

File No. 0206170-000 January 2023

HALEY & ALDRICH, INC.

2033 N. Main Street Suite 309 Walnut Creek, CA 94596 925.949.1012 www.haleyaldrich.com

17 January 2023 File No. 0206170-000

900 South Broadway Denver, Colorado 80209

Attention: Aicha Menendez

Subject: Design Level Geotechnical Investigation Report

Stinson Beach Lifeguard Tower and Visitor Snack Bar Replacement Marin County, California

Ladies and Gentlemen:

Enclosed is our draft geotechnical investigation report for the proposed Stinson Beach Lifeguard Tower and Visitor Snack Bar Replacement in Stinson Beach, California.

We understand the proposed project will include the construction of a new three-story Lifeguard Tower and Visitor Snack Bar building with a gross footprint of 2,800 square feet. The new building will include a covered deck, an at-grade parking garage, a wooden boardwalk, a wood retaining wall, and a concrete slab-on-grade. Additional improvements include underground utilities and other site improvements, including new pavements and hardscapes as well as landscape features.

This report contains discussions regarding subsurface conditions, seismicity and seismic hazards, and foundation recommendations. We find that the primary geotechnical issues that should be addressed during the design and construction of the planned project includes strong seismic shaking, potentially liquefiable soil layers up to 50 feet below ground surface, and undocumented fill. To mitigate these potential issues, we recommend that the proposed three-story building be supported on a deep foundation system of prestressed concrete piles that penetrates through the fill and potentially liquefiable soil layers to gain support in underlying native alluvial clays and sands. The deck and boardwalk will be supported on timber piles.

Without mitigation, settlement following a design level earthquake is estimated to be on the order of 6 to 8 inches total and up to 4 to 6 inches of differential settlement over 50 feet or between columns.

With mitigation of deep foundations, total settlement at the columns following a design level seismic event is anticipated to be less than 1 inch with ½ inch of differential settlement between the columns.

This will not, however, protect unimproved areas such as slab-on-grade or exterior improvements.

Our recommendations regarding foundations, Site grading, fill compaction, and other geotechnical aspects of this project are presented in this geotechnical investigation report. The recommendations contained in this report are based on a limited subsurface investigation. Variations between expected

17 January 2023 and actual soil conditions may be found in localized areas during construction. Therefore, we should be retained to observe Site preparation, foundation installation, and placement of fill and backfill, during which time we may make changes to our recommendations if deemed necessary.

We appreciate the opportunity to provide our services to you on this project. If you have any questions, please call Catherine at 925.949.4407 or 971.277.8945.

Sincerely yours, HALEY & ALDRICH, INC.

Michael Liu, PE (WA) Staff Professional, Geotechnical

Catherine Ellis, PE*, GE** Senior Associate, Geotechnical Engineer

*CA, OR, WA; **CA

Reviewed By:

Christopher R. Nardi, PE, GE Senior Technical Expert, Geotechnical Engineer

Enclosures

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Table of Contents Page i

List of Tables iii List of Figures iii List of Appendices iv

1. Introduction 1

1.1 PURPOSE AND SCOPE 1

2. Site Conditions and Proposed Construction 2

2.1 PROJECT DESCRIPTION 2

3. Geology 3

3.1 REGIONAL GEOLOGY 3

3.2 REGIONAL SEISMICITY 3

4. Subsurface Explorations 6

4.1 CONE PENETRATION TESTS 6

4.2 FIELD INVESTIGATION BY OTHERS 7

4.3 LABORATORY TESTING 7

5. Subsurface Soil and Groundwater Conditions 8

5.1 SUBSURFACE SOIL CONDITIONS 8

5.2 SHEAR WAVE VELOCITY CONDITIONS 8

5.3 GROUNDWATER CONDITIONS 8

6. Discussion and Conclusions 9

6.1 SEISMICITY AND SEISMIC HAZARDS 9

6.1.1 Site Seismicity 9

6.1.2 Soil Liquefaction 10

6.1.3 Cyclic Densification 11

6.1.4 Lateral Spreading 11

6.1.5 Sand Boils 11

6.1.6 Tsunami 11

6.1.7 Fault Rupture 12

6.1.8 Expansion Potential 12

6.2 FOUNDATIONS AND SETTLEMENT 12

7. Recommendations 13

Table of Contents Page ii

EARTHWORKS 13

7.1.1 Site Preparation and Grading 13

7.1.2 Subgrade Preparation 13

7.1.3 Material for Fill 14

7.1.4 Reuse of On-Site Material 14

7.1.5 Imported, Non-Expansive Fill 15

7.1.6 Fill Placement and Compaction 15

7.1.7 Weak or Wet Subgrade Mitigation 16

7.1.8 Underground Utilities 16

7.2 PRECAST, PRESTRESSED CONCRETE PILES 17

7.2.1 Axial Pile Capacity 17

7.2.2 Lateral Resistance of Piles 17

7.2.3 Lateral Capacities 18

7.2.4 Indicator Pile Program 19

7.3 WOODEN PILES FOR BOARDWALKS AND DECK 19

7.3.1 Lateral Resistance of Piles 19

7.3.2 Lateral Capacities 19

7.3.3 Indicator Pile Program 20

7.4 SLABS-ON-GRADE 20

7.4.1 Floor Slabs 20

7.4.2 Exterior Flatwork 21

7.5 RETAINING WALL 21

7.6 UNDERGROUND UTILITIES 22

7.7 FLEXIBLE PAVEMENT DESIGN 22

7.8 SURFACE DRAINAGE 23

7.9 WET WEATHER CONSTRUCTION 23

7.10 CORROSION POTENTIAL 24

8. Supplemental Geotechnical Services 25

9. Limitations 26

References 27 iii

List of Tables

Table No. Title

I Active Faults within 100 km of the Site (embedded, p. 3)

II Seismic Design Parameters (embedded, p. 9)

III Values Used in Liquefaction Evaluation (embedded, p. 10)

IV Estimated Liquefaction Settlement (embedded, p. 11)

V General Engineered Fill Grading Requirements (embedded, p. 14)

VI Class 2 Aggregate Base Grading Requirements (embedded, p. 14)

VII Imported Fill Requirements (embedded, p. 15)

VIII Summary of Compaction Recommendations (embedded, p. 15)

IX LPILE Parameters for Precast Concrete Piles (embedded, p. 18)

X Timber Pile Properties (embedded, p. 19)

XI Gradation Requirements for Capillary Moisture Break (embedded, p. 22)

XII Flexible Pavement Design (Subgrade R-Value = 33) (embedded, p. 23)

List of Figures

Figure No. Title

1 Project Locus

2 Site Plan

3 Regional Geologic Map iv

List of Appendices

Appendix Title

A Cone Penetration Test Results

B Previous Investigations by Others

C Laboratory Testing

D Piles Design Charts

1. Introduction

Stinson Beach, located in Marin County, California, along the California coast, is part of the National Park System. Many of the built structures in this area date to the 1950s and have had limited improvement up to the present day. This report presents the results of our geotechnical investigation of the proposed Stinson Beach Lifeguard Tower and Visitor Snack Bar replacement project located at the Golden Gate National Recreation Area (GGNRA) in Stinson Beach, Marin County, California (Site), as shown on Figure 1, “Project Locus” and Figure 2, “Site Plan.” This report describes the subsurface explorations performed at the Site, provides our interpretation of the subsurface data, and includes geotechnical recommendations for the proposed development.

1.1 PURPOSE AND SCOPE

The scope of our services was presented in our revised proposal dated 4 August 2022. Our investigation proposal included the advancement of three cone penetration test (CPT) soundings, including one deep seismic CPT, to perform engineering analyses to develop conclusions and recommendations regarding:

soil and groundwater conditions at the Site;

Site seismicity and seismic hazards, including estimates of liquefaction potential, cyclic densification, and lateral spreading, if appropriate;

measures to mitigate potential seismic hazards, if any;

measures to improve existing fill, if appropriate;

appropriate foundation type(s) for the proposed building;

foundation design criteria, including vertical and lateral capacities;

estimates of foundation settlement;

expansion potential of the on-Site soils based on geotechnical laboratory tests and provision of proper mitigation methods, if appropriate;

design criteria for ground floor slabs;

Site grading, including criteria for fill quality and compaction;

flexible and rigid pavement designs;

seismic design parameters in accordance with the 2019 California Building Code (CBC); and construction considerations.

2. Site Conditions and Proposed Construction

The Site is located in the western portion of Stinson Beach in Marin County, California, and is situated along the edge of the shoreline at approximate coordinates of 37.8959° N and 122.6395° W. The Site is included within the GGNRA and is currently occupied by the wooden lifeguard tower and the old Siren Canteen beachside grill. These facilities are accessed by an asphalt driveway and parking lot located north of the amenities. The Site is bounded by the Pacific Ocean and the active shoreline to the west, vegetation on the southeast and northwest sides, and an asphalt paved driveway on the northeast side of the building that leads to the parking lot. The elevation of the asphalt driveway adjacent to the building is 25 feet (per the Site Layout Plan prepared by DHM Design dated 5 December 2022), which decreases to the northeast with an average grade of 9 percent until the driveway changes direction. The driveway then gradually slopes towards the parking lot. The northern end of the Site includes a hill with an elevation of 30 feet at its highest point and is the maximum elevation within the Site. The lowest elevation at the Site is 19 feet and is located at the base of the hill.

2.1 PROJECT DESCRIPTION

It is our understanding that the proposed project will include the demolition of the existing lifeguard tower and grill structure, and the construction of a new Lifeguard Tower and Visitor Snack Bar building totaling three stories and a gross area of 2,800 square feet. The new building will include a partially covered deck, an at-grade parking garage, a wooden boardwalk, a wood retaining wall, and a concrete slab-on-grade. The project will also include improvement of underground utilities and other Site improvements, including new pavements and hardscapes as well as landscape features.

Based on our communications with the Structural Engineering team, we understand that maximum column loads will be on the order of 70 kips of dead plus live load. We have assumed wall loads on the order of 2 to 3 kips per linear foot.

Grading for the Site will be limited to cuts and fills on the order of 2 to 3 feet. Wood retaining walls up to 6 feet are proposed. Trenches for utilities have been assumed to be 5 feet or less in depth.

If the project differs significantly from the conditions described above, we should be consulted to review the applicability of our recommendations.

3. Geology

3.1 REGIONAL GEOLOGY

The Site is located within the Coast Ranges Geomorphic Province of California. Throughout the Cenozoic Era, the western part of California has been affected by tectonic forces associated with lateral or transform plate motion between the North American and Pacific crustal plates, which has produced a complex system of northwest-trending faults - the San Andreas, Hayward, and Calaveras Fault systems being the most prominent within the Bay Area. Uplift, erosion, and subsequent redeposition of sedimentary rocks within this province have been driven primarily by the northwest-southeast directed strike-slip movement of the tectonic plates and the associated northeast-oriented compressional stress.

The northwest-trending coastal mountain ranges are the result of an orogeny believed to have been occurring since the Pleistocene epoch (approximately 2 to 3 million years before present). A Regional Geologic Map of the Site is presented in Figure 3.

The Coast Ranges Geomorphic Province is geologically separated into a series of tectonostratigraphic terranes that are northwest-oriented and fault-bounded due to the prevailing tectonic influence of the San Andreas fault system and crustal plate boundary on the area. There are generally two types of basement rocks within the region; older, highly deformed rocks that have been moved long distances by large-scale fault displacement and younger, less deformed rocks of the Tertiary Period that are relatively closer to their source of deposition (Blake et al., 1984).

The Site is situated along a spit of sand within the Bolinas Lagoon littoral cell. Littoral cells are closed systems of deposition along the coast, typically constrained between two coastal prominences or rock outcrops. North of the Site, the Duxbury Reef prominence is underlain by Santa Cruz Mudstone (Miocene) and Rocky Point; south of Stinson Beach comprises Franciscan Complex Mélange (Cretaceous). Between these two points, dune and beach sands of the Quaternary Period have accreted and formed the spit that underlies the Stinson Beach shoreline (Blake et al., 2000).

3.2 REGIONAL SEISMICITY

Coastal Northern California is a region of high seismicity, owing to the tectonics of the crustal plate boundary and the San Andreas Fault System. The major Holocene-active faults in the area are presented in Table I. For each of the Holocene-active faults within 100 kilometers (km) of the site, the distance, fault characteristics, and estimated maximum Moment magnitude, Mw, are summarized in Table I (2014 Working Group on California Earthquake Probabilities, 2015).

TABLE I

Active Faults within 100 km of the Site

Fault Name Distance

(km)

Direction from Site

Mean Characteristic or

Maximum Moment

Magnitude

Mean

Slip Rate

(mm/yr)

Fault

Length

(km)

San Andreas - 1906 Rupture 2 West 7.9 19 473

San Andreas - North Coast South 2 West 7.5 24 191

San Andreas - Peninsula 14 Southeast 7.2 17 85

Northern San Gregorio 14 South 7.2 7 110

Fault Name Distance

(km)

Direction from Site

Mean Characteristic or

Maximum Moment

Magnitude

Mean

Slip Rate

(mm/yr)

Fault

Length

(km)

Total San Gregorio 14 South 7.4 5 176

Point Reyes 18 West 6.8 0.3 47

North Hayward 27 Northeast 6.5 9 35

Total Hayward 27 Northeast 6.9 9 88

Total Hayward-Rodgers Creek 27 Northeast 7.3 9 151

Rodgers Creek 28 Northeast 7.0 9 63

South Hayward 38 East 6.7 9 53

West Napa 46 Northeast 6.5 1 30

Concord/Green Valley 52 East 6.7 -- 56

Mt Diablo - MTD 53 East 6.7 2 25

Total Calaveras 56 East 6.9 -- 123

Monte Vista-Shannon 62 Southeast 6.8 0.4 41

Greenville 71 East 6.9 2 51

Hunting Creek-Berryessa 73 Northeast 6.9 6 60

Maacama-Garberville 76 North 6.9 9 182

Great Valley 6 76 East 6.7 1.5 45

Great Valley 5 79 East 6.5 1.5 28

Great Valley 4 79 Northeast 6.6 1.5 42

Hayward - Southeast Extension 81 Southeast 6.4 3 26

San Andreas - Santa Cruz Mtns. 98 Southeast 7.0 17 62

Collayomi 98 North 6.5 0.6 29

Great Valley 7 100 East 6.7 1.5 45

Notes:

km = kilometer(s) mm/yr = millimeters per year

Since 1800, four major earthquakes have been recorded on the San Andreas Fault. In 1836, an earthquake with an estimated maximum intensity of VII on the Modified Mercalli (MM) scale occurred east of Monterey Bay, reportedly on the San Andreas Fault (Toppozada and Borchardt, 1998). The estimated Moment magnitude, Mw for this earthquake was about 6.25. This earthquake was previously thought to have occurred on the northern portion of the Hayward Fault. In 1838, an earthquake occurred on the San Andreas Fault with an estimated intensity of about VIII-IX (MM), corresponding to an Mw of about 7.5. The San Francisco Earthquake of 1906 caused the most significant damage in the history of the Bay Area in terms of loss of lives and property damage. This earthquake created a surface rupture approximately 470 km in length along the San Andreas Fault, from Shelter Cove to San Juan Bautista. It had a maximum intensity of XI (MM), an Mw of about 7.9, and was felt 560 km away in Oregon, Nevada, and Los Angeles. The Loma Prieta Earthquake of 17 October 1989 occurred on the San Andreas Fault in the Santa Cruz Mountains. It had an Mw of 6.9 and was approximately 33 km from the Site.

In 1868, an earthquake with an estimated maximum intensity of X on the MM scale occurred on the southern segment (between San Leandro and Fremont) of the Hayward Fault. The estimated Mw for the earthquake was 7.0. In 1861, an earthquake of unknown magnitude (probably an Mw of about 6.5) was reported on the Calaveras Fault. The most recent significant earthquake on this fault was the 1984 Morgan Hill earthquake (Mw = 6.2).

The most recent major earthquake to affect the Bay Area was the West Napa Earthquake of 24 August 2014. The West Napa Earthquake had an Mw of 6.0 and was located approximately 86 km from the Site.

The third Uniform California Earthquake Rupture Forecast (UCERF3) prepared by the U.S. Geological Survey (USGS) reports a 72 percent probability of a magnitude 6.7 or greater earthquake occurring in the San Francisco Bay region (which includes the Site) by the year 2044 (2014 Working Group on California Earthquake Probabilities, 2015). More specifically, the 30-year probabilities of a magnitude

6.7 or greater earthquake are 1.2 percent for the Monterey Bay-Tularcitos Fault, 1.4 percent for the San Gregorio Connected Fault, and 15.4 percent for the San Andreas Fault.

4. Subsurface Explorations

Haley & Aldrich, Inc. (Haley & Aldrich) based the findings presented in this report on our 2022 field investigation and the two recent geotechnical investigations by Yeh and Associates (2020) and Kleinfelder (2016) that were made available to us. Haley & Aldrich explored subsurface conditions at the Site by advancing three CPT soundings and one shallow infiltration test on 15 November 2022. Prior to performing our field investigation, we contacted Underground Service Alert (USA) and retained a private utility locator to clear the CPT locations of existing utilities. The approximate locations and designations of the subsurface explorations are shown on Figure 2, Site Plan. A summary of the various exploration programs is presented below.

4.1 CONE PENETRATION TESTS

On 15 November 2022, three (3) CPTs (designated HA-CPT-1 through HA-CPT-3) were advanced to depths between 53.81 and 97.93 feet below ground surface (bgs) by Middle Earth Geo Testing, Inc.

(Middle Earth) of Hayward, California. The upper 5 feet of soil material at each CPT location was hand augered, and our field representative logged the generated spoils. The CPTs encountered refusal at depths above the planned 100-foot depth. Pore pressure dissipation tests were performed in HA-CPT-1 and HA-CPT-3 to measure hydrostatic water pressures and infer the approximate depth to groundwater.

The CPTs consisted of hydraulically pushing a 1.75-inch-diameter, cone-tipped probe into the soil using a rig with a push capacity of 20 tons. The cone tip measures tip resistance, and the friction sleeve behind the cone tip measures frictional resistance. Electrical strain gauges within the cone continuously measured soil parameters during the entire depth the cone was advanced. Soil data, including tip resistance and frictional resistance, were recorded and then processed to provide information for use in our geotechnical engineering analyses. Once completed, the CPT holes were backfilled with cement grout in accordance with the Marin County Health Department Division of Environmental Health permit requirements.

To conduct the seismic shear wave test, the penetration of the cone is stopped and the rods are decoupled from the rig. An automatic hammer is triggered to send a shear wave into the soil. The distance from the source to the cone is calculated knowing the total depth of the cone and the horizontal offset distance between the source and the cone. Seismic shear wave tests were performed at every 5-foot interval to a depth of 85 feet at HA-CPT-3.

The stratigraphic interpretation of the CPT data was performed based on relationships between cone bearing and sleeve friction versus penetration depth. The friction ratio, which is sleeve friction divided by cone bearing, is a calculated parameter used to infer soil behavior type. Generally, cohesive soils (clays) have high friction ratios, low cone bearing, and generate large excess pore water pressures.

Cohesionless soils (sands) have lower friction ratios, high cone bearing, and generate small excess pore water pressures. The interpretation of soil properties from the cone data was carried out using recent correlations developed by Robertson and Cabal (2010). The CPT logs and shear wave velocity tests generated for the exploration program by Middle Earth, showing tip resistance and friction ratio by depth and interpreted soil classifications and strengths, are presented in Appendix A.

4.2 FIELD INVESTIGATION BY OTHERS

Two other geotechnical investigations were previously performed near the Site. One was conducted by Yeh and Associates (dated 22 June 2020) for the GGNRA Improvements Project, which included improvements at Stinson Beach, including rehabilitation of parking lots, access roads, drainage and flood control and improving Americans with Disabilities Act (ADA) access. The Site investigation consisted of seven hollow stem auger borings to a depth of 5 feet bgs and one to a depth of 50 feet bgs.

Groundwater was encountered at approximately 1 foot bgs and 4 feet bgs during drilling. The subsurface data from the Yeh and Associates report are presented in Appendix B.

A geotechnical investigation was previously performed by Kleinfelder (dated 29 July 2016) for an ADA-compliant beach access walkway, stairway, and concrete pad located at the Stinson Beach restroom facility. The investigation consisted of one hand-augered hole advanced to about 5½ feet bgs and three Dynamic Cone Penetrometer tests to depths between about 15 and 16 feet bgs. Groundwater was measured at approximately 4.5 feet bgs in the hand-augered boring. The Site plan and boring logs are presented in Appendix B.

4.3 LABORATORY TESTING

Soil samples were collected by Haley & Aldrich from soil cuttings and transported to Inspection Services, Inc. in Berkeley, California, for geotechnical laboratory testing. The samples were tested for gradation and for Resistance (R)-Value. Additional near-surface soil samples were submitted to CERCO Analytical, Inc. in Concord, California, and tested for corrosion properties, including pH, resistivity, sulfate content, and chloride content. The laboratory test results are presented in Appendix C.

5. Subsurface Soil and Groundwater Conditions

5.1 SUBSURFACE SOIL CONDITIONS

The results of the current and past subsurface investigations indicate that the Site is underlain by alluvial materials including predominantly clayey and sandy soils as indicated on the CPT logs advanced during our investigation and included in Appendix A. The boring logs from the previous consultants’ investigations are included in Appendix B. The shallow fills at the Site are underlain by alluvial overburden consisting of stiff silty clays and medium dense to dense silty sands, sands, and dense gravelly sand soils. The results from the three CPTs advanced during our investigation were relatively consistent and indicated a subsurface profile consisting of stiff, thin, silty clay layers interbedded with silty sands to a depth of approximately 25 feet bgs. Medium dense silty sand and medium stiff sandy silt were encountered to a depth of approximately 70 feet. Finally, stiff clayey silts and fine-grained soils thinly interbedded with sand were encountered to our deepest exploration depth of about 98 feet bgs.

5.2 SHEAR WAVE VELOCITY CONDITIONS

A shear wave velocity profile of subsurface soils at the Site was recorded by means of seismic CPT shear wave testing at HA-CPT-1 and -3. Since HA-CPT-1 encountered refusal at about 54 feet, we based our evaluation on the data from HA-CPT-3 which extended to about 98 feet. The test results indicate that the average shear wave velocity of the soils within the upper 100 feet (30 meters) of the Site is approximately 166 meters per second (543 feet per second [ft/s]).

5.3 GROUNDWATER CONDITIONS

The depth to groundwater at the Site was estimated using groundwater level measurements from the pore pressure dissipation test data collected at the CPT locations. Pore pressure dissipation (PPD) tests used to estimate the depth to groundwater at each location were performed at HA-CPTs 1-3 on 15 November 22. Based on the PPD test results, we estimate that groundwater at the time of testing was present at depths between 5 and 8 feet below the existing surface grade, which corresponds to approximate elevations ranging from 12 to 15 feet (NAVD 881). As noted in the two studies by others where groundwater was encountered higher, between 1 and 4½ feet below grade, it can fluctuate across the Site. Groundwater levels can fluctuate based on seasonal rainfall amounts, perched groundwater conditions, and elevation changes in nearby bodies of water including tidal changes near the active shoreline. For planning and design purposes, a stabilized groundwater level at elevation 15 feet should be anticipated at the Site.

1 North American Vertical Datum of 1988 (NAVD).

6. Discussion and Conclusions

On the basis of our review of subsurface information for the Site, we conclude that the proposed project is feasible from a geotechnical standpoint. The primary geotechnical issues that should be addressed during the design and construction of this project include:

Site seismicity and seismic hazards, including the potential for soil liquefaction and liquefaction-induced settlement;

The presence of undocumented fill; and the selection of an appropriate foundation system that will support the planned structure during static and seismic loading conditions.

Our discussion of these and other issues are presented in the remainder of this report.

6.1 SEISMICITY AND SEISMIC HAZARDS

During a major earthquake, seismic shaking has the potential to occur at the Site, as is typical throughout the Bay Area, and as experienced during the 1989 Loma Prieta event. Shaking during an earthquake can result in ground failure, such as that associated with soil liquefaction, lateral spreading, and cyclic densification. Haley & Aldrich’s assessment of these potential seismic hazards are presented in the following sections.

6.1.1 Site Seismicity

A seismic hazard analysis was performed using the USGS Unified Hazard Tool website (USGS, 2021). Our deaggregation analysis utilized the USGS Dynamic Conterminous U.S. 2014 (v.4.2) edition. For our analyses, we evaluated the Site as Site Class D, based on an average shear wave velocity (Vs30) over the top 80 feet (26 meters) of the Site of about 543 ft/s; this value was calculated using the average shear wave velocity measured at sounding HA-CPT-2 during our field investigation on 15 November 2022.

Based on the seismicity of faults that may impact the Site and the results of the deaggregation analysis, a design earthquake with a mean Mw of 7.7 was selected for the seismic hazard evaluation. The peak horizontal ground acceleration (PHGA) for the Site, which is based on the Maximum Considered Earthquake (MCE) with a return interval of 2,475 years, or a 2 percent probability of exceedance in 50 years, is 0.881g. The risk-based site-modified peak ground acceleration (PGAM) for the Site is 1.101g;

this value was computed based on procedures outlined in ASCE 7-16 (ASCE, 2017). Recommended code-based seismic parameters for designing the proposed structures in conformance with the 2019 California Building Code (CBC, 2019) and ASCE 7-16 are presented in Table II.

TABLE II

Seismic Design Parameters

Seismic Parameter Design Value1

Site Class (ASCE 7-16 Table 1613.5.2) D

Risk Category II

Seismic Design Category E2

Seismic Parameter Design Value1

MCER

3 Ground Motion (Period = 0.2 seconds), Ss 2.341

MCER Ground Motion (Period = 1.0 seconds), S1 0.979

Peak Ground Acceleration, PGA 1.001g4

Site Amplification Factor at 0.2 seconds, Fa 1.2

Site Amplification Factor at 1.0 seconds, Fv --5

Site Amplification Factor for PGA, FPGA 1.1

Site-Modified Peak Ground Acceleration, PGAM 1.101 g

Site-Modified Spectral Acceleration Value at 0.2 seconds, SMS 2.341

Site-Modified Spectral Acceleration Value at 1.0 seconds, SM1 --5

Design Spectral Acceleration at 0.2 seconds, SDS 1.56

Design Spectral Acceleration at 1.0 seconds, SD1 --5

Notes:

1) Design values presented above are based on a site located at latitude / longitude = 37.89902 / -122.64552.

2) E corresponds to buildings of Occupancy Groups I, II and III in areas NEAR MAJOR ACTIVE FAULTS. Soil or rock is of no consequence.

3) MCER = Risk-targeted maximum considered earthquake

4) g = acceleration of gravity

5) Values of Fv, SM1, and SD1 are undefined for this site class without performance of a site-specific ground motion hazard analysis.

6.1.2 Soil Liquefaction

Liquefaction is the process in which saturated, cohesionless soil experiences a temporary loss of strength based on the buildup of excess pore water pressure during cyclic loading resulting from earthquake ground motions. The type of soils most susceptible to liquefaction are loose, clean, saturated, uniformly graded sand and silt that have low clay content. Flow failure, lateral spreading, differential settlement, loss of bearing strength, ground fissures, and sand boils are evidence of liquefaction.

Depths to groundwater at the Site during our investigation were about 5 to 8 feet bgs in November 2022. A depth of 5 feet was used in the analysis based on the conditions encountered in the field. We evaluated liquefaction potential at the Site by performing analysis using the software CLiq in accordance with the methodology by Boulanger & Idriss (2014). The parameters used in the liquefaction evaluation are shown in Table III.

TABLE III

Values Used in Liquefaction Evaluation

Liquefaction Evaluation Parameter Value

Depth to Groundwater, during current explorations (feet bgs) 5-8

Depth to Groundwater, during current explorations (elevation, NAVD 88) 12 to 15

Depth to Groundwater, during Design Earthquake (feet bgs) 5

Design Peak Ground Acceleration 1.101g

Predominant Earthquake Moment Magnitude, Mw 7.8

Factor of Safety for Liquefaction Triggering 1.3

Based on our current geotechnical exploration at the proposed lifeguard station, we estimated that liquefaction settlement will be generally 6 to 8 inches and conclude that the potential for on-Site liquefaction to occur within the upper 50 feet bgs and adversely impact the planned structures can be high.

TABLE IV

Estimated Liquefaction Settlement

Exploration No.

Estimated Liquefaction

Settlement Upper 50 feet (inches)

HA-CPT-1 6

HA-CPT-2 7½

HA-CPT-3 8

6.1.3 Cyclic Densification

Seismically-induced compaction or densification of non-saturated granular soil (such as sand above the groundwater table) based on earthquake vibrations can result in settlement of the ground surface.

Based on the results of our subsurface exploration program, we estimate that settlement from cyclic densification will be less than ¼-inch based on the high groundwater table, the settlement is included in the values summarized in Table IV.

6.1.4 Lateral Spreading

Lateral spreading is a potential hazard commonly associated with liquefaction where extensional ground cracking and settlement occur as a response to lateral migration of subsurface liquefiable material.

These phenomena typically occur adjacent to free faces such as slopes and creek channels. The Youd et

al. study (2001) was developed using an empirical methodology for estimating the lateral spread horizontal displacement using a database where the depth from the ground surface to the top of the liquefiable layer is limited to 30 feet. Using that methodology, and the results of our subsurface exploration program, we conclude that the potential for lateral spreading at the Site to be high. The use of a pile foundation system should reduce the potential for damage to the structure.

6.1.5 Sand Boils

Based on Ishihara (1985), we believe that the potential for ground surface disruption (such as sand boils, ground fissures, etc.) to occur at this Site is high. Another major concern during an earthquake is some form of ground surface disturbance or ground failure. The ground failure can be in the form of sand boils, small ground fissures, ground oscillation, such as buckled pavements, curbs, broken pipelines, etc., and lateral ground displacement. One of the major reasons for ground surface disruption is insufficient cover thickness of a non-liquefiable layer over a liquefiable layer. Given the relatively thin cap of non-liquefiable soil above the potentially liquefiable layer, we anticipate the potential for ground surface disruption to be high.

6.1.6 Tsunami

The State of California has produced a series of maps illustrating coastal zones that are at risk of inundation during a tsunami event. The Site in its entirety is located on the “County of Marin, Tsunami Hazard Area Map” (Patton and Wilson, 2022). The entire Stinson Beach spit, including the Site, is located within a tsunami hazard area. Therefore, we judge that the potential for a seismically induced wave to impact the Site to be high. The highlands surrounding the spit on either side are located outside of the tsunami hazard zone.

6.1.7 Fault Rupture

Historically, ground surface displacements closely follow the trace of geologically young faults. The Site is not within an Earthquake Fault Zone, as defined by the Alquist-Priolo Earthquake Fault Zoning Act.

Therefore, we conclude the risk of surface faulting and secondary ground failure to be low.

6.1.8 Expansion Potential

Expansive soils are characterized by their ability to undergo significant volume change (shrink or swell) based on variations in moisture content. Changes in soil moisture content can result from rainfall, landscape irrigation, perched groundwater, drought, or other factors. Changes in soil moisture may result in unacceptable settlement or heave of structures, concrete slabs supported on-grade, or pavements supported on these materials. Given the sandy nature of the soils, the potential for expansion conditions is very low.

6.2 FOUNDATIONS AND SETTLEMENT

Fill materials primarily consisting of clay with silt, sand, and gravel were encountered within the upper 1 to 3 feet of the Site. Fill and medium dense native alluvium within about 50 feet bgs are potentially liquefiable. The results of our investigation indicate that the clayey alluvial soils increase in strength with depth. If the proposed three-story building is supported on shallow foundations that bear on the relatively weak and compressible fill and alluvial soils, the building will likely experience excessive and erratic settlement resulting in cracks and damage to the building.

To reduce the potential for building settlement, the proposed building loads should be supported on a deep foundation system that will transfer the loads to competent alluvial soils lying below the fill and weak liquefiable alluvial silt and clay deposits. Based on our qualitative evaluation of various deep foundation systems, we conclude that driven, 14-inch-square, precast, prestressed concrete piles would likely be the most economical foundation system for supporting the proposed building.

Other deep foundation systems such as drilled piers, full-displacement auger cast piles, and torque-down piles were considered for use at this Site. However, based on the downdrag load and cost considerations, we conclude that a pile rig driving precast piles into the underlying soil would be the most efficient method of installing deep foundations at the Site.

As an alternative, we considered using ground improvement to stiffen and strengthen the fill layer and the underlying weak and liquefiable alluvial soil layers. However, in our opinion, the cost of treating/improving the existing fill and native alluvium can become expensive in order to improve the potential liquefiable soil to 50 feet bgs.

7. Recommendations

Our geotechnical recommendations for the proposed building, as well as other Site improvements, are presented in this section of the report.

EARTHWORKS

7.1.1 Site Preparation and Grading

The proposed building and parking lot areas should be cleared of existing pavements, trees, abandoned utilities, and other obstructions. The proposed building and parking lot areas should be stripped of soil containing over 3 percent organic matter (if present). We anticipate the excavation for this project can be made using conventional earth-moving equipment.

Existing utilities within the proposed building’s footprint should be completely removed. Existing utilities within the parking lot areas should be completely removed where appearing at depths of less than 4 feet below the proposed ground surface elevation. Where existing utilities within proposed parking lot areas are deeper than 4 feet below the proposed ground surface, the utilities may either be removed, as described above, or abandoned in place by exposing, cutting, and capping the ends of the lines, and placing cement grout within the lines to fill existing voids. We recommend that abandoned-in-place lines be grouted in 100-foot linear sections to confirm that grout has fully filled the abandoned pipe.

7.1.2 Subgrade Preparation

The Site should be rough-graded to accommodate the proposed grading plan. Recommendations for mitigating existing fill are presented for each building type below. In proposed building areas, subgrade preparation should extend at least 5 feet beyond the limits of the proposed building slabs and any adjoining flatwork. In exterior concrete slab and pavement areas, subgrade preparation should extend at least 2 feet beyond the limits of these improvements.

We estimate that approximately 6 feet of undocumented fill is present in the vicinity of the proposed building. Within the building’s footprint, loose or disturbed soil or undocumented fill soils should be over-excavated to a minimum of 2 feet below the planned bottom of the slab. The exposed subgrade should be scarified to a depth of at least 12 inches, moisture-conditioned, and compacted in accordance with the recommendations given in the section entitled “Fill Placement and Compaction.”

The on-Site soils can become disturbed under construction traffic, especially if the soils are wet. If soils become disturbed, they should be over-excavated to the underlying undisturbed soils. The excavated soil can then be dried and recompacted back into place, or it could be removed and replaced with drier import fill. The exposed subgrade should be scarified to a depth of at least 12 inches, moisture-conditioned, and compacted in accordance with the recommendations given in the section entitled “Fill Placement and Compaction.”

In non-foundation areas that will receive new fills, building loads, and Site improvements, such as pavements, sidewalks, and slabs, the exposed soil subgrade should be evaluated by the Geotechnical Engineer. If it is fill, it should be over-excavated to a minimum of 1 foot below existing grade or finished pad elevation, whichever is deeper. The exposed subgrade should be scarified to a depth of at least

12 inches, moisture-conditioned, and compacted in accordance with the recommendations given in the section entitled “Fill Placement and Compaction.” If it is undisturbed, competent native soil, consideration may be given by the Geotechnical Engineer to reduce the over-excavation.

Prepared soil subgrades should be non-yielding when proof-rolled by a fully loaded water truck or equipment of similar weight. Moisture-conditioning subgrade soil should consist of adding water if the soil is too dry and allowing the soil to dry if the soil is too wet. After the subgrades are properly prepared, the areas may be raised to design grades by placing engineered fill.

7.1.3 Material for Fill

Except for organic laden soil, the on-Site soil is suitable for use as general engineered fill if it is free of deleterious matter and satisfies the criteria in Table V. Soil for use in engineered fill should be inorganic, and free of deleterious materials and hazardous substances. For this project, inorganic soil is soil with an organic content of less than 3 percent by weight or without visible organic matter deemed excessive by Haley & Aldrich. Site recycled material may be processed and reused as engineered fill if it meets the requirements presented in this report for the specific materials.

TABLE V

General Engineered Fill Grading Requirements

Sieve Size Percentage Passing Sieve

3-inch 100

1½-inch 85-100

7.1.4 Reuse of On-Site Material

Any existing asphalt or aggregate base that is removed during demolition may be suitable to be pulverized and mixed with the underlying base for use as engineered fill if it has an organic content of less than 2 percent by dry weight and meets the requirements presented under the “Material for Fill” section of this report.

The processed asphalt concrete/base material may be used as Class 2 Aggregate base if it meets the following requirements from Section 26 of the Caltrans Standard Specifications:

TABLE VI

Class 2 Aggregate Base Grading Requirements

Sieve Size Percentage Passing

1-inch 100 min.

¾-inch 90-100

No. 4 35-60

No. 30 10-30

No. 200 2-9

Note Quality Requirements:

Sand Equivalent: 25 min R-value: 78 min

7.1.5 Imported, Non-Expansive Fill

All imported fill soils should be nearly free of organic or other deleterious debris, essentially non-plastic, and less than 3 inches in maximum dimension. Specific requirements for import fill are provided below.

TABLE VII

Imported Fill Requirements

Sieve Size Percentage Passing Sieve

3-inch 100

1½-inch 85-100

#200 Screen 8-40

Atterberg Limits Percent

Plasticity Index 12 or less

Liquid Limit Less than 30

Fill materials should be approved by the project Geotechnical Engineer prior to placement and delivery to the Site. At least 5 working days prior to importing, a representative sample of the proposed import fill should be delivered to us for laboratory for evaluation.

7.1.6 Fill Placement and Compaction

Fill materials should be placed and compacted in horizontal lifts, each not exceeding 8 inches in uncompacted thickness. Compaction of fill should be performed by mechanical means only. Based on equipment limitations, thinner lifts may be necessary to achieve the recommended degree of compaction. Fill should be placed in accordance with Table VIII, Summary of Compaction Recommendations.

TABLE VIII

Summary of Compaction Recommendations

Area Compaction Recommendations (See Notes 1 through 4)

Subgrade Preparation and Placement of General Engineered Fill,5 Including Imported Non-expansive Fill

Compact upper 12 inches of subgrade and entire fill to a minimum of 90 percent relative compaction at near to slightly over optimum moisture content. Where interior flatwork is exposed to vehicular traffic, compact aggregate base to a minimum of 95 percent compaction at near optimum moisture content.

Trenches6 Compact trench backfill to a minimum of 90 percent relative compaction at near to slightly over optimum moisture. Where trenches will be under the pavement section, flatwork, or other improvements, the upper 12 inches, measured from finished grade of the trench backfill should be compacted to a minimum of 95 percent compaction at near optimum moisture content.

Exterior Flatwork Compact upper 12 inches of subgrade to a minimum of 90 percent relative compaction at near to slightly over optimum moisture content. Compact aggregate base to a minimum of 90 percent compaction at or above optimum moisture content. Where exterior flatwork is exposed to vehicular traffic, compact aggregate base to a minimum of 95 percent compaction at near optimum moisture content.

Area Compaction Recommendations (See Notes 1 through 4)

Paved Areas Compact upper 12 inches of subgrade to a minimum of 95 percent relative compaction at near to slightly over optimum moisture content. Compact aggregate base rock to a minimum of 95 percent relative compaction at near optimum moisture content.

Notes:

1) Depths are below finished subgrade elevation.

2) All compaction requirements refer to relative compaction as a percentage of the laboratory standard described by ASTM D-1557

(latest version). All lifts to be compacted shall be a maximum of 8 inches loose thickness.

3) All compacted surfaces, such as fills, subgrades, and backfills need to be firm and stable, and should be unyielding under compaction equipment.

4) Where fills, such as backfill placement after removal of existing underground utility lines, are greater than 7 feet in depth, the portion of the fill deeper than 7 feet should be compacted to a minimum of 95 percent compaction.

5) Includes building pads.

6) In landscaping areas, this percent compaction in trenches may be reduced to 85 percent. Water jetting or flooding to obtain compaction of backfill should not be permitted.

7.1.7 Weak or Wet Subgrade Mitigation

The depths to groundwater were at around 5 and 8 feet bgs during our subsurface explorations in mid- November 2022 and may fluctuate over time. Excavations for foundations, utilities, and other improvements may encounter weak or wet soil conditions depending on the depths and elevations of these features. If weak or wet soil subgrade is encountered during grading and adequate compaction cannot be achieved, the Contracting Officer and Geotechnical Engineer should be notified immediately to assess the condition of the weak or wet subgrade and provide Site-specific recommendations for stabilizing and/or repairing the exposed subgrade. Potential subgrade repair options include:

Over-excavating and removing the weak or wet soil and replacing it with select non-expansive fill underlain by geotextile tensile fabric (Mirafi 500X or equivalent).

Stabilizing the exposed subgrade by thoroughly blending a lime or cement admixture into the weak or wet soil at a concentration of approximately 5 percent of dry weight of soil being treated, and subsequently compacting the treated subgrade to at least 90 percent relative compaction.

7.1.8 Underground Utilities

Excavations for utility trenches can be made with a backhoe. All trenches should conform to the current Cal/OSHA requirements. Backfill for utility trenches and other excavations is also considered fill, and it should be compacted according to the recommendations presented in the section “Fill Placement and Compaction.” Jetting of trench backfill should not be permitted. Special care should be taken when backfilling utility trenches in pavement areas. Poor compaction may cause excessive settlements, resulting in damage to the pavement section.

Underground utilities should be located above a 1.5:1 (horizontal to vertical) plane projected downward from the bottom of the new footings to avoid undermining the footings during the excavation of the utility trench.

Pipes or conduits should be supported on bedding material with a thickness equal to D/4 (with D equal to the outside diameter of the pipe) or 4 inches of sand or fine non-angular gravel below the pipe, whichever is greater. After the pipes and conduits are tested, inspected (if required), and approved, they should be covered to a minimum depth of 6 inches with sand or fine non-angular gravel, which should be mechanically tamped.

Soil excavated from utility trenches may require moisture-conditioning (drying) prior to reusing the material as compacted on-Site general fill. Although not anticipated, excavations extending below groundwater will require dewatering. Dewatering should lower the groundwater level to a minimum of 2 feet below the bottom of the excavation. If the soil exposed in the bottom of the excavation is soft or wet, it will be necessary to over-excavate the soil and replace it with crushed rock to create a working platform. The depth of over-excavation should be determined in the field at the time of construction;

however, for planning purposes, a depth of 12 inches may be assumed.

7.2 PRECAST, PRESTRESSED CONCRETE PILES

The proposed building may be supported on driven concrete piles that gain support through skin friction in the alluvial soil deposits that underlie the existing fill.

7.2.1 Axial Pile Capacity

In our opinion, a driven 14-inch-square, precast, prestressed concrete pile that gains support through a combination of skin friction and end bearing is the most appropriate type of pile to support the proposed building. Support from the upper 50 feet of potentially liquefiable soil layer of fill and alluvial soil should be ignored during compression and seismic uplift conditions.

We estimate that an approximately 75-foot-long pile that is driven to a tip of elevation -55 feet will have an ultimate dead plus live load capacity of 550 kips as shown in Figure D-1 in Appendix D. With a factor of safety of 2.0 applied, the allowable static load would be 275 kips.

Following procedures from Ishihara, we considered the effects of liquefaction on the piles (Ishihara, 1995). As a result, with the downdrag from liquefaction, allowable total load capacity (including seismic and wind loads) of the piles should be limited to 100 kips of ultimate load. Applying a factor of safety of 1.5, the total allowable load pile capacity is 67 kips. No reduction factor has been applied to downdrag forces. Figure D-1 shows the ultimate compression resistance in Appendix D.

For uplift resistance in static condition, a 75-foot-long, 14-inch-square concrete pile will have an ultimate capacity of 55 kips, or an allowable uplift capacity of 18 kips; this value has a factor of safety of

3.0. Figure D-1 in Appendix D shows the ultimate uplift resistance.

The piles should be spaced no closer than three pile widths, center to center, to avoid reductions to the capacities due to group effects. An indicator pile program and pile testing should be performed to verify the capacities of the piles and the factors of safety used in the pile design.

7.2.2 Lateral Resistance of Piles

Lateral load resistance can be mobilized by the individual piles in combination with other foundation elements embedded below the ground surface.