Geotechnical Report.pdf

NP ROCR 15(2)

Joyce Road Bridge at Military Road NW Reconstruction

Final Geotechnical Engineering Report

October 2023

Eastern Federal Lands Highway Division

Draft Pre lim inary Ge otec hnical Engi nee ring Re por t Easter n Fed era l Lan ds Highway Dist rict

Final Geotechnical Engineering Report

NP ROCR 15(2) – Joyce Road Bridge at Military Road NW Reconstruction

Project No: W3X71281

Document Title: Geotechnical Engineering Report

Document No.: n/a

Revision: Final Design

Document Status: n/a

Date: October 17, 2023

Client Name: Eastern Federal Lands Highway District

Client No: NP ROCR 15(2)

Project Manager: Chris Naab

Author: Phillip Gilmore

File Name: Final GER - Joyce Road Bridge.docx

Jacobs Engineering Inc.

2551 Dulles View Drive, Suite 700 Herndon, Virginia 20171 United States T +1.703.817.4800 www.jacobs.com

© Copyright 2019 Jacobs Engineering Inc. The concepts and information contained in this document are the property of Jacobs. Use or copying of this document in whole or in part without the written permission of Jacobs constitutes an infringement of copyright.

Limitation: This document has been prepared on behalf of, and for the exclusive use of Jacobs’ client, and is subject to, and issued in accordance with, the provisions of the contract between Jacobs and the client. Jacobs accepts no liability or responsibility whatsoever for, or in respect of, any use of, or reliance upon, this document by any third party.

Document history and status

Revision Date Description Author Checked Reviewed Approved

Draft 9/18/2020 Preliminary Geotechnical Engineering Report PG KC EF

Draft 3/8/2021 Phase III Geotechnical Engineering Report PG KC EF

Draft 6/11/2021 Phase IV Geotechnical Engineering Report PG ES EF

Final 10/24/2023 Final Geotechnical Engineering Report PG CN EF

Contents

1. Introduction 5

1.1 Project Description 5

1.2 Purpose and Scope 5

1.3 Assumptions and Design Criteria 5

2. Geology and Site Conditions 7

2.1 Regional Geology 7

2.2 Site Geology 7

2.3 Site Conditions 8

3. Subsurface Conditions 10

3.1 Historic Information 10

3.2 Subsurface Investigation 10

3.3 Laboratory Testing Program 11

3.3.1 Soil Classification and Index Testing Results 11

3.3.2 Soil Compaction and Testing Results 12

3.3.3 Corrosivity Testing Results 12

3.3.4 Rock Strength Testing Results 13

3.4 Stratigraphy 14

3.4.1 Stratum 1: Fill Material 14

3.4.2 Stratum 2: Residual Soil 14

3.4.3 Stratum 3: Bedrock 14

3.5 Design Soil Profile and Parameters 14

3.6 Groundwater Conditions 16

4. Geotechnical Recommendations 17

4.1 Foundation Type Selection 17

4.2 Design Loads and Design Parameters 17

4.3 Resistance and Load Factors 18

4.4 Scour Considerations 19

4.5 Micropile Foundation Design 21

4.5.1 Geometry and Micropile Layout 21

4.5.2 Axial and Structural Analysis 22

4.5.3 Lateral Analysis 22

4.5.4 Global Stability Analysis 24

4.6 Foundation Recommendations 25

4.7 Erosion Protection 25

4.8 Pavement Recommendations 26

5. Construction Considerations 27

5.1 Excavation for Structure 27

5.1.1 Groundwater and Surface Water Control 27

5.2 Micropile Installation 27

5.2.1 Load Verification Test 28

5.2.2 Load Proof Test 28

6. Limitations 29

7. References 30

Appendix A. Relevant Historic Drawing with Subsurface Information

Appendix B. Boring Location Plan

Appendix C. Rock Core and Pavement Core Photos

Appendix D. Borehole Logs

Appendix E. Report of Laboratory Results

Appendix F. Geotechnical Analyses and Results

Acronyms and Abbreviations

AASHTO American Association of State Highway and Transportation Officials ASTM American Society for Testing and Materials bpf blows per foot CBR California Bearing Ratio EFLHD Eastern Federal Land Highway Division FHWA Federal Highway Administration GEC Geotechnical Engineering Circular GER Geotechnical Engineering Report ksi kips per square inch LRFD Load Resistance Factors Design NPS National Park Service pcf pounds per cubic foot PDDM Project Development and Design Manual ppm parts per million psi pounds per square inch psf pounds per square foot RMR rock mass rating RQD rock quality designation SPT standard penetration test STS split tensile strength TGM Technical Guidance Manual USGS United States Geological Survey UCS unconfined compressive strength WSDOT Washington State Department of Transportation

1. Introduction

The Eastern Federal Land Highway Division (EFLHD) of the Federal Highway Administration (FHWA), along with its partner agency National Park Service (NPS), contracted Jacobs for the engineering services required for reconstruction of Joyce Road Bridge (Structure No. 3450-003P) at Military Road NW, formally called Project ROCR 15(2). This Geotechnical Engineering Report (GER) presents the findings of a subsurface investigation and provides foundation recommendations for the proposed Joyce Road Bridge replacement.

1.1 Project Description

The Joyce Road Bridge is a single-span concrete tee beam bridge located in Rock Creek Park, Washington D.C. The bridge was built in 1950 and is currently in poor condition due to deterioration of the structural concrete (15-ton weight limit), cracking in the asphalt wearing surface, and failed joints.

In March 2014, as part of the Bridge Inspection and Management Program, FHWA inspected the bridge and attempted strength tests on concrete cores, however, tests were not completed due to core deterioration. The concrete cores strength test failure and deterioration to the deck and superstructure resulted in poor rating condition of the most critical elements.

The general objective of Project ROCR 15(2) is to replace the superstructure with precast concrete beams and a cast-in-place concrete deck. It is anticipated that the new stub abutments will be supported on deep foundations and the new abutments will be constructed immediately behind the existing abutments to support the superstructure. The existing abutment retaining walls are anticipated to remain in place.

1.2 Purpose and Scope

The purpose of this geotechnical report is to present a summary of collected geotechnical data and provide foundation recommendations for the new Joyce Road Bridge. The scope of this report included:

● Reviewing available geological publications and historical geotechnical information

● Conducting a subsurface exploration program consisting of two boreholes

● Evaluating exploratory borehole data and laboratory testing data

● Developing geotechnical design and construction recommendations

1.3 Assumptions and Design Criteria

The analyses presented in this report were developed based on the following assumptions and limitations:

The suitability of the foundation type recommended in this report is based on the proposed alignment and available cross sections.

Recommendations are presented based on the latest construction staging, scheduling, and maintenance of traffic plans. These recommendations will need to be reviewed and possibly modified if changes are made to those plans.

The following design standards were used in the analyses and the development of geotechnical recommendations:

American Association of State Highway and Transportation Officials (AASHTO) Load Resistance Factors Design (LRFD) Bridge Design Specifications, 8th Edition, 2017. Hereinafter referred to as the AASHTO LRFD specifications.

FHWA Federal Lands Highway Project Development and Design Manual (PDDM), Chapter 6, July 2018.

FHWA Geotechnical Engineering Circular (GEC) No. 5 – Geotechnical Site Characterization, FHWA- NHI-16-072, April 2017.

FHWA Geotechnical Technical Guidance Manual (TGM), May 2007.

FHWA Reference Manual, Micropile Design and Construction, FHWA-NHI-05-039, December

2005.

WSDOT Geotechnical Design Manual, Washington State Department of Transportation, M 46-

03.09, December 2013.

The foundations are designed in accordance with the AASHTO LRFD approach. Foundations are evaluated against the critical load combinations for service limit state, strength limit states, and extreme limit state.

Global stability of the abutment was evaluated using the service limit state based on Service I Load Combination and an appropriate resistance factor. The loading and resistance factors used in the analyses are discussed further in Section 4.3. The lateral deflection of the foundation cap is limited to 0.5 inches in service limit state and 1.0 inch in strength limit states with the exception of the 100-yr scour case. The foundations are designed for a capacity-to-demand ratio (CDR) greater than 1.0 across all critical load combinations and design conditions.

2. Geology and Site Conditions

2.1 Regional Geology

The project site is in Rock Creek Park which lies within the Piedmont physiographic province and contains metamorphosed sedimentary and igneous rock of the Cambrian to Ordovician age. This includes granodiorite, gneiss, tonalite, schist, and serpentinitic rocks. The rocks have been folded, deformed, and recrystallized during three separate tectonic events: the Taconic, Acadian, and Alleghanian orogenies. In some areas, these rocks are covered with a mantle of relatively young rock debris, alluvium, regolith, and slope deposits (NPS, 2009).

Rock Creek Park regionally marks the Fall Zone, a transitional zone where harder, more resilient metamorphic rocks to the west intersect the softer, less consolidated sedimentary rocks of the Coastal Plain to the east, to form an area of ridges and waterfalls. The nearly vertical fault zone cuts the bedrock and trends north to south just west of Rock Creek. This fault was active during the igneous intrusive event and then reactivated during the late Paleozoic Alleghanian Orgogeny. Following the last major uplift phase, the Piedmont was eroded into a plateau. Surficial deposits eroded from the mountains to the west combined with the marine deposits from the east to form the Atlantic Coastal Plain (NPS, 2009).

2.2 Site Geology

The project site is located in the eastern part of Rock Creek Park, which is underlain predominantly by the Laurel Formation of Lower Cambrian age (see Figure 2-1). This formation generally consists of medium-to coarse-grained, moderately to well-foliated metamorphic rocks, including meta-arenite, biotite schist, actinolite schist, amphibolite, and—at a greater degree of foliation—gneiss. The rocks are generally dark brown, gray to dark gray, and olive in color. The overlying soils generally consist of a weathered (saprolite) zone of the Piedmont Formation. These soils are the result of in-place weathering of the underlying rock.

Alluvial deposits consisting of clean sands and gravels exist within and around the nearby Luzon Branch.

Figure 2-1: Site Geology – Adapted from the Geologic Map of the Washington West Quadrangle, District of Columbia, Montgomery and Prince Georges Counties, Maryland, and Arlington and Fairfax Counties, Virginia

2.3 Site Conditions

The Joyce Road Bridge site is located in Rock Creek Park, one of the largest natural urban parks in the country. Joyce Road is a two-lane road with an alignment that generally runs north-south (see Figure 2-2).

The Joyce Road Bridge overpasses Luzon Branch, a small stream that generally flows parallel with Joyce Road from north to south and into Rock Creek. The bridge abutments and adjacent retaining walls provide grade separation between Joyce Road and the lower stream channel (see Figure 2-3). To the west of the site is the Rock Creek Golf Course consisting of open, grassy fields with a rolling hill to flat terrain.

Immediately south of the site, Military Road NW overpasses Joyce Road. East of the site is 16th Street NW (oriented north-south) plus housing and business development which is separated from the site by a small woodland area surrounding Luzon Branch.

Legend of Map Units

Alluvium of stream bottoms and flood plains; swamp deposits; and tidal deposits (Holocene)

Gravel, sand, silt, and clay (late Miocene)

Leucocratic granitoids, undifferentiated (Early Ordovician)

Laurel Formation (Lower Cambrian) (Hopson, 1964)

Potomac Group (Lower Cretaceous)

Site

Figure 2-2: Existing site conditions of Joyce Road Bridge and Military Road overpass viewed from road surface (viewing southward)

Figure 2-3: Existing site conditions of Joyce Road Bridge and retaining walls viewed from Luzon Branch (viewing westward)

3. Subsurface Conditions

3.1 Historic Information

Two historical soil boreholes were performed as part of the design and construction of the existing Joyce Road Bridge. The historical subsurface information was referenced from historic construction drawings prepared by the Department of Highways, D.C. Bridge Design Section, Office of Planning, Design, and Engineering (n.d.). The relevant historic drawing showing the borehole locations and borehole logs is presented in Appendix A. Note that the elevations shown in the as-built historic drawings are likely to be based on NGVD 29.

3.2 Subsurface Investigation

The recent subsurface investigation was performed by Froehling & Robertson, Inc. (F&R) in June 2020 under the direct supervision of Jacobs. The investigation consisted of two (2) geotechnical boreholes. A summary of drill depths and elevation data for completed boreholes is presented in Table 3-1. A borehole location plan is provided in Appendix B.

Table 3-1 – Borehole Summary

Borehole ID Top of Borehole Elevation (ft)1

Total Depth of Borehole (ft)2

Depth in Soil (ft)

Depth in Rock

(ft) Location

20BH-BR-01 176.7 50.5 21.5 29.0 South Abutment

20BH-BR-02 181.0 60.3 36.0 24.3 North Abutment

Notes:

1 The elevations are based on NAVD 88.

2 Depths measured from existing Joyce Road ground surface.

The boreholes were advanced using a CME-75 Truck‐Mounted drilling rig. A pavement core was collected at each borehole. The boreholes were drilled with 3‐1/4‐inch inside-diameter (ID) hollow‐stem augers (HSA) in general accordance with ASTM D1452. Standard Penetration Tests (SPTs) were performed using a standard 2-inch outside-diameter (OD) split-spoon sampler driven 18 or 24 inches with a 140-pound automatic hammer in general accordance with ASTM D1586. Rock coring was performed using NQ-sized diamond fitted onto a double tube swivel-type core barrel.

The number of blows required to drive the sampler each 6-inch interval was recorded, where the sum of SPT blow counts for the second and third 6-inch intervals represents the SPT N-value as defined by ASTM D1586. The field recorded SPT N‐values are not corrected for hammer efficiency. Therefore, the N60 value (N‐value adjusted to a 60 percent hammer efficiency) can be obtained by multiplying the field recorded SPT N‐value by the energy correction factor, which is defined as the efficiency of the hammer divided by 60 percent. The energy correction factor for the automatic hammer used in this exploration is 1.48.

Jacobs geotechnical engineer logged the boreholes in the field in general accordance with ASTM D5434 and AASHTO standards. Soil and rock core samples were collected for visual classification and laboratory testing. Disturbed 24-inch split-spoon samples were retrieved continuously the upper 10 feet and at maximum intervals of 5 feet thereafter. Rock quality designation (RQD) and total core recovery were recorded in inches and percentages for each retrieved rock core run. Rock core and pavement core photos are provided in Appendix C.

Additionally, one (1) bulk bag sample was collected from 20BH-BR-01 between 1.5 to 6.5 feet below ground surface (bgs) for additional laboratory testing. Two (2) grab samples were collected from the stream bed of Luzon Branch; samples were collected approximately 50 to 60 feet upstream of Joyce Road Bridge. A third pavement core was also collected at an offset location 20BH-BR-02A to further evaluate the existing pavement conditions farther north of the north end of the existing bridge.

Groundwater levels in the boreholes were recorded at the time of drilling and upon completion of drilling.

During drilling, the groundwater was recorded if encountered prior to introduction of water to advance the boreholes especially in rock. After completion, boreholes were cement grouted to top of rock and soil cuttings were discarded in the borehole to bottom of pavement and patched with cold set asphalt.

Borehole logs were prepared in general accordance with EFLHD guidelines, referencing the FHWA Geotechnical Technical Guidance Manual (TGM) and Project Development and Design Manual (PDDM). It is noted that some variations can be expected between visually classified and laboratory‐classified samples; however, the differences are minor and are not expected to affect engineering evaluations.

Borehole logs of this subsurface exploration are presented in Appendix D.

3.3 Laboratory Testing Program

A laboratory testing program as summarized in Table 3-2 was performed by F&R on select soil samples to verify field classifications and to determine engineering properties. A letter report by F&R on the laboratory test results is presented in Appendix E of this report. The test results are further summarized in subsequent sections.

Table 3-2 –Laboratory Testing Program

Test Standard Number of Tests

Sieve analysis (w/o hydrometer) ASTM D6913 6

Hydrometer ASTM D4221 1

Atterberg Limits ASTM D4318-05 3

Moisture Content ASTM D2216-05 8

Percentage Passing No. 200 ASTM C117-90 4

Corrosivity: pH, Sulfate, Chloride, and Resistivity 1 1 1

Standard Proctor Compaction Testing AASHTO T99 1

California Bearing Ratio (3-point) AASHTO T193 1

Unconfined Compressive Strength and Elastic Moduli of Intact Rock Core Specimens

ASTM D7012,

Method D

Unconfined Compressive Strength ASTM D7012 Method C 4

Splitting Tensile Strength of Intact Rock Core Specimens ASTM D3967-08 3

Notes:

1 ASTM D4972, ASTM C1580, AASHTO 291, AASHTO T288.

3.3.1 Soil Classification and Index Testing Results

Select soil samples were tested in the laboratory for moisture content, Atterberg limits, and grain size distribution to determine soil classification and index properties of the soil. The results of the testing are summarized in Table 3-3 and detailed results presented in Appendix E.

Table 3-3 – Summary of Soil Classification and Index Properties

Borehole ID Sample Depth

Moisture Content

LL PL PI % Gravel % Sand % Fines USCS AASHTO

20BH-BR-01 1.5-3.0 8.7 30 20 10 21.0 54.8 24.1 SC A-2-4

20BH-BR-01 5.0-7.0 14.3 11.5 33.9 54.6

20BH-BR-01 9.0-11.0 22.8 31 30 1 2.8 74.9 22.3 SM A-2-4

20BH-BR-01 13.0-15.0 13.8 4.5 75.6 19.9

20BH-BR-02 3.2-5.2 10.5 26.7 52.4 20.8

20BH-BR-02 7.2-9.2 12.2 28 17 11 34.6 34.9 30.5 SC A-2-6

20BH-BR-02 13.0-15.0 18.7 77.7 22.3

20BH-BR-02 18.0-20.0 14.6

20BH-BR-02 28.0-30.0 12.0 0.4 77.1 22.5

SB-11 1.0 61.4 38.4 0.2 GP

SB-21 3.0 50.7 48.0 1.3 GW

Notes:

1 Samples SB-1 and SB-2 were collected from the streambed of Luzon Branch at 1.0 and 3.0 feet, respectively.

3.3.2 Soil Compaction and Testing Results

A Standard Proctor compaction test was performed in general accordance with AASHTO T99-15 Method A Standard on a bulk sample collected from 1.5 to 6.5 feet in 20BH-BR-01. A 3-point California Bearing Ration (CBR) test was performed in general accordance with AASHTO T193. The results of the Proctor compaction and CBR testing are summarized in Table 3-4 and detailed results presented in Appendix E.

Table 3-4 – Summary of Compaction and CBR Testing

Borehole ID

USCS

Classification

AASHTO

Classification Optimum Moisture

Content (%)

Maximum Dry Density

(pcf)

Swell

CBR Value1

20BH-BR-01 SC A-2-4 9.0 129 0.0 13.5

Notes:

1 CBR Value at 95% of Maximum Dry Density based on penetration at 0.1 in.

3.3.3 Corrosivity Testing Results

A suite of corrosivity tests including pH, resistivity, sulfate content, and chloride content was performed on selected samples from 20BH-BR-02. Testing was performed in accordance with respective ASTM or AASHTO standards as listed in Table 3-2. A composite sample combining soils from split-spoon samples SS-3 to SS-10 was used for testing. The results of the corrosivity tests are summarized in Table 3-5 and detailed results presented in Appendix E.

Table 3-5 – Summary of Corrosivity Testing

Borehole ID Sample Depth

(ft) pH Minimum Resistivity (ohm-cm)

Sulfate Content (mg/Kg)

Chloride Content (mg/Kg)

20BH-BR-02 5.17 – 33.251 5.6 6,508 < 10 56

Notes:

1 Testing was performed on a composite sample using split-spoon samples SS-3 to SS-10.

Acid sulfate soils may cause degradation of construction materials, such as steel and concrete, and weathering of fill material and volumetric changes detrimental to pavement. Consequently, consideration of the effects of corrosion and deterioration from environmental conditions is required in the selection of the pile type. The following soil or site conditions should be considered as indicative of a potential corrosion situation:

Resistivity less than 2,000 ohm-centimeters pH less than 5.5 Sulfate concentrations greater than 1,000 parts per million

Following suggestions in FHWA GEC No. 5 (2017), the testing indicates moderately corrosive environment for steel based on moderately high resistivity values. The low sulfate content indicates soils are generally noncorrosive to concrete.

3.3.4 Rock Strength Testing Results

Unconfined compressive strength (UCS) testing and split tensile strength (STS) testing were performed on selected rock core samples in accordance with ASTM D7012 (Method C or D) and ASTM D3967, respectively. Test results including unconfined compressive strength, modulus of intact rock (Eint), and split tensile strength are summarized in Table 3-6 and detailed results presented in Appendix E.

Table 3-6 – Summary of Rock Strength Testing

Borehole ID Core Run

Sample Depth (ft)

RQD (%)

UCS

(psi)

Eint

(ksi)

STS

(psi)

20BH-BR-01 R-1 25.5 53 8,581

20BH-BR-01 R-2 27.36 – 27.46 100 1,630

20BH-BR-01 R-4 33.0 88 7,700

20BH-BR-01 R-4 33.33-33.71 88 18,9062 16,500

20BH-BR-01 R-5 38.95 – 39.04 100 1,240

20BH-BR-02 R-1 39.07 – 39.17 53 1301

20BH-BR-02 R-2 44.00-44.38 68 23,2402 11,300

20BH-BR-02 R-2 44.7 68 12,784

20BH-BR-02 R-4 54.2 90 11,803

20BH-BR-02 R-4 55.50-55.88 90 17,6012 13,000

Notes:

1 Specimen within highly weathered/weak rock zone did not meet size requirements for UCS testing.

2 Peak compressive stress report in the table.

3 Young’s modulus at the relevant stress range is specified in the table.

3.4 Stratigraphy

The site stratigraphy at Joyce Road Bridge was evaluated based on borehole data and collected soil samples during the subsurface exploration. The subsurface was generally separated into three distinct strata: Fill material, Residual soil, and Bedrock. The subsequent sections present the general soil classifications, density/consistency, SPT N-values, and general thickness of each stratum.

3.4.1 Stratum 1: Fill Material

Fill material potentially placed during the construction of the existing Joyce Road Bridge was encountered to depths of about 9 and 11 feet bgs in 20BH-BR-01 and 20BH-BR-02, respectively. The fill was generally composed of brown, clayey sand and sandy clay with varied amounts of gravel. SPT N-values within the fill generally varied from 7 to 15 blows per foot (bpf) with an average of about 10 bpf.

3.4.2 Stratum 2: Residual Soil

Residual soils formed by in-place weathering of the underlying bedrock were encountered immediately below fill material. The identification of the residual soil was apparent by the relic rock structure observed in the soil samples. The residual soils consisted of brown silty sand. This stratum can be divided into two sublayers: an upper loose to medium dense layer and a lower very dense intermediate geomaterial (IGM) layer. The upper residual soils (5 to 7 feet thick as encountered in 20BH-BR-01 and 20BH-BR-02, respectively) had SPT N-values of 6 to 19 bpf. The lower IGM layer generally had SPT N-values greater than 50 blows per 6 inches. The IGM was encountered at elevation 163 feet in both borings, having a thickness of 9 to 18 feet in 20BH-BR-01 and 20BH-BR-02, respectively. The increased thickness at 20BH- BR-02 is likely due to the pre-existing location of Luzon Branch under the north abutment of the existing Joyce Road Bridge.

3.4.3 Stratum 3: Bedrock

Bedrock was encountered at 21.5 and 36.0 feet bgs in 20BH-BR-01 and 20BH-BR-02, respectively. The bedrock mostly consists of dark gray, medium grained, moderately foliated schist. The rock varied from highly weathered to slightly weathered, and moderately to slightly fractured with zones of highly fractured rock. A zone of dark brown, highly weathered and weak schist was encountered from 38.0 to 42.5 feet in 20BH-BR-02. The RQD values varied from 21 to 100 percent, and core recovery from 68 to 100 percent, across both boreholes and generally improved with depth.

3.5 Design Soil Profile and Parameters

The findings of subsurface investigation were used to develop site-specific soil and rock parameters for the three strata encountered at the project site. Some engineering properties of the soil were directly measured by laboratory or in situ soil testing, whereas other properties were interpreted based on published correlations with SPT N-values and index properties. Methodology and correlations presented in FHWA GEC No. 5 (2017), Liang (2002), and Kulhawy and Mayne (1990) were used to develop the ranges of soil parameters presented in Table 3-7.

Table 3-7 – Summary of Correlated Soil Parameters

Stratum Bottom

Elevation N60

(bpf)

(pcf)

(deg) E

(ksf) k

(pci)

Existing Fill 168 10-22 (15)

120-137 (124)

31-33 (32)

110-152 (131)

60-90 (75)

Residual 163 9-28 (19)

119-128 (124)

30-36 (33)

101-207 (462)

50-135 (93)

IGM 154 to 145 73-100

(97) 137-140

(140) 49-57 (54)

420-544 (530)

330-445 (432)

Notes:

The range of values represent the minimum and maximum values. The values shown in parentheses () represent average values.

Recommended soil parameters for geotechnical design were deduced from the above correlated values, available laboratory testing results, and engineering judgement. These design soil parameters, as presented in Table 3-8, were used to carry out the engineering evaluation of the proposed abutment foundations and proposed temporary excavations.

Table 3-8 – Summary of Recommended Design Soil Parameters

Stratum Soil Type

(pcf)

(deg) E

(ksf) k

(pci)1

Fill Clayey Sand 125 30 125 75

Residual Silty Sand 125 30 150 90 (45)

IGM Silty Sand 135 45 500 400 (200)

Notes:

1 Subgrade reaction for saturated soils are in parentheses ().

The design parameters for rock were evaluated based on laboratory testing results and correlations presented in AASHTO LRFD Bridge Design Specifications (2017). Since the bedrock will primarily account for the geotechnical capacity of the micropiles, specific parameters were developed at each abutment location based on the encountered condition of the rock and corresponding rock testing results. The rock mass parameters used in the foundation analyses are presented in Table 3-9.

Table 3-9 – Summary of Recommended Design Rock Parameters

Rock Parameter Strong Rock

(Abutment 1) Weak Rock2

(Abutment 2) Strong Rock2

(Abutment 2)

Unit Weight (pcf) 145 145 145

Unconfined Compressive Strength (psi) 8,600 780 11,250

Split Tensile Strength (psi) 1,435 130 1,435

Geologic Strength Index, GSI1 60 15 50

Material Index 10 10 10

Poisson’s Ratio 0.3 0.3 0.3

Intact Modulus (ksi) 16,500 11,300 12,150

Mass Modulus (ksi) 1,980 40 1,270

Notes:

1 GSI was estimated using the minimum of two methods: (1) correlated from Rock Mass Rating (RMR) after Hoek and Brown (1997) and (2) correlated based on joint condition rating and RQD after Hoek, Carter and Diederichs (2013).

2 The bedrock encountered in 20BH-BR-02 (Abutment 2) consisted of an upper highly weathered layer that was considerably weaker than the underlying bedrock, therefore, rock parameters were developed for two layers: weak rock and strong rock.

3.6 Groundwater Conditions

Artesian groundwater conditions were observed at 20BH-BR-01 after rock coring was advanced to approximately 44 feet bgs. Once encountered, water was observed flowing up and out at the ground surface throughout the progression of the borehole and at completion on June 11, 2020, as shown in Figure 3-1. The borehole was backfilled and patched with asphalt at the surface, however, water was still observed seeping out from the patched borehole on June 12, 2020. The borehole was re-drilled the same day, backfilled with a denser cement aggregate grout, and re-patched with asphalt. Water seepage was not observed on subsequent days after re-backfilling and re-patching. Nearby well data referenced from a groundwater assessment study for the District of Columbia (DC WRRC Report No. 127, 1993) cited similar semi-artesian conditions in the weathered zone underlying the Rock Creek Golf Course.

Groundwater was also encountered at about 13 feet bgs in 20BH-BR-02 during drilling. Artesian groundwater conditions were not observed at 20BH-BR-02.

Figure 3-1: Observed water flowing out of 20BH-BR-01 after encountered on June 11, 2020 (left) and after re-drilling the backfilled borehole on June 12, 2020 (right).

4. Geotechnical Recommendations

4.1 Foundation Type Selection

Soils encountered in the upper 11 feet at proposed locations of the abutments predominantly consisted of fill or residual soils with a variable composition and strength. For the design loads at the proposed abutments, large shallow foundations will be required to satisfy the tolerable settlement limits. Moreover, lateral loads imposed by the proposed structure will not be adequately resisted by soil at relatively shallow depths. One option is to excavate and replace the loose soils with compacted suitable soil. However, given the depth of the fill and residual deposits, groundwater condition, available right-of-way, existing abutment walls and existing structures adjacent to the proposed bridge elements, excavation and replacement of the loose fill and residual soils is not considered a cost-effective alternative. Therefore, deep foundations are recommended for the abutments supporting the bridge.

The new bridge superstructure will be supported by stub abutments on deep foundations located behind the existing bridge abutments. Due to site constraints, proposed loading, and soil conditions, micropile foundations are recommended to support the new bridge. Micropiles are a very versatile deep foundation option and can be installed in a variety of difficult underground environments such as those presented by manmade fills with obstructions, cobbles and boulders, bedrock, karstic terrain, and artesian groundwater conditions.

The proposed foundation pile caps will be located behind the existing abutment foundations; two proposed micropiles will be installed through the existing wall foundation at Abutment 1. The proposed micropiles will also be installed into bedrock. Based on the findings of the subsurface investigation, the encountered rock quality generally improved with increased depth, therefore the bedrock is expected to provide considerable bearing resistance for the proposed structure.

Micropiles provide further flexibility for the site constraints posed by the overhead clearance of Military Road Bridge and the existing abutment and retaining wall foundations. Micropile rigs are generally small and can be easily lifted by cranes to access an installation area that is confined to an excavation. This is beneficial to the project due to site disturbance constraints and required excavation at both abutments;

the deep foundation installation at the site will require excavation depths of 7 to 8 feet to bottom of cap elevations. Installation also causes minimal vibration, therefore reducing the potential for impact to the existing walls and neighboring bridge.

4.2 Design Loads and Design Parameters

The general design parameters used in the foundation evaluation are summarized in Table 4-1.

Unfactored loads imposed by the bridge superstructure on the bearing pedestals and abutments are presented in Appendix F.

Table 4-1 – Bridge Foundation Design Parameters

Structure Abutment 1 (South) Abutment 2 (North)

General Design Parameters

Corresponding borings 20BH-BR-01 20BH-BR-02

Approximate finished grade elevation 1 177 feet 181 feet

Approximate top elevation of bedrock 1 154 feet 145 feet

Micropile Design Parameters

Steel casing size 2 10.75-inch diameter and 0.514-inch thick

Steel reinforcing bar #14 threaded bar

Cut-off elevation 1 171.5 feet 175.5 feet

Notes:

1 Elevations are approximated.

2 A sacrificial thickness of 0.0625 inches was used in the analyses to account for the possible effects of corrosion to the steel casing.

4.3 Resistance and Load Factors

The resistance and load factors summarized in Table 4-2 were used in the micropile foundation design, in accordance with AASHTO LRFD specifications (2017).

Table 4-2 – Resistance and Load Factors Used in Micropile Analyses

Resistance and Load Factors Value

Side resistance in rock for micropile 0.55

Tip resistance in rock for micropile (O’Neill and Reese, 1999) 0.50

Global stability resistance factor 0.65

Micropile lateral resistance factor 1.00

Structural resistance factor of cased length in compression 0.75

Structural resistance factor of uncased length in compression 0.75

Structural resistance factor of cased length in tension 0.80

Structural resistance factor of uncased length in tension 0.80

Resistance factor for concrete in shear 0.90

Resistance factor for steel member in shear 1.00

Resistance factor for axial compression – combined axial and flexural resistance 0.80

Resistance factor for flexure of steel – combined axial and flexural resistance 1.00

Horizontal at rest earth pressure load factor (EH) 1.35

Horizontal active earth pressure load factor (EH) 1.50

Vertical earth pressure load factor for overall failure—Strength I (EV) 1.00

DC component load factor—Strength I (DC) 1.25

Vertical earth pressure surcharge load factor—Strength I (ES) 1.50

Live Load surcharge load factor—Strength I (LL) 1.75

Load combination limit states evaluated in the analyses, and general load factors used according to Table 3.4.1-1 in the AASHTO LRFD specifications (2017), are summarized in Table 4-3.

Table 4-3 – Load Combinations and Load Factors

Load Combination Limit State 1

Load Factors

DC LL WS WL TU EH

Service I 1.00 1.00 1.00 1.00 1.20 1.00

Strength I 1.25 1.75 0.00 0.00 1.20 1.35

Strength III 1.25 0.00 1.00 0.00 1.20 1.35

Strength V 1.25 1.35 1.00 1.00 1.20 1.35

Extreme II 1.00 0.50 0.00 0.00 0.00 1.00

Notes:

1 Changes in foundation conditions resulting from design flood for scour were considered at strength and service limit states. Changes due to scour resulting from the check flood were considered at the extreme event limit state.

4.4 Scour Considerations

Due to the existing stream, the consequences of longitudinal scour along channel walls were accounted for in the foundation analyses. The longitudinal scour depth along the existing walls is estimated to be 6 feet for the 100-year design flood and 7 feet for the 200-year check flood. The micropile foundations were designed according to the criteria in Section 1.3 while also considering the scour condition for the 100-year design flood in service and strength limit states and the 200-year check flood for extreme event limit state, in accordance with AASHTO LRFD specifications (2017).

The depth of scour was conservatively estimated from the bottom of the existing footing (elevation 163 feet) at the south abutment due to exposed footings observed along the south wall in its current condition. Approximately 4 feet of sediment build-up was observed above the existing footing at the north abutment, therefore, depth to scour was estimated from the top of sediment at approximate elevation 169 feet. The profile of the scour hole used in the evaluation of the micropiles is estimated assuming a 1H:1V slope within IGM and 2H:1V slope in residual and fill soils; refer to the illustration in Figure 4-1 for estimated profile at the south abutment wall.

Based on the presented scour profiles for 100-year and 200-year floods, it is expected complete failure of the existing retaining walls would occur in all service, strength, and extreme limit states. Given the current state of scour at the retaining wall foundations, and no known historical issues due to scour, it is unlikely that scour of this degree will occur and cause complete failure of the existing retaining wall. However, the micropile foundations were designed for this worst-case scenario, and assumes all live loads and load factors following AASHTO LRFD specifications (2017) are applicable—i.e., the bridge is still in service despite failure of the existing retaining wall. Overall, the scour condition controlled the micropile design.

Figure 4-1: Estimated scour profile at existing south abutment wall footing.

Due to limitations of the software used for the lateral analysis—unable to model a sloped ground surface—the depth of scour used in analyses was estimated as the scour profile elevation located approximately 4 times the pile diameter from the micropiles; refer to Figure 4-1. The corresponding scour elevations presented in Table 4-4 were used in the analyses.

Table 4-4 – Scour Elevations Used in Analyses

Abutment Bottom of

Existing Footing Elevation

Top of Stream Bedding Elevation

Scour Elevation for Design Flood

Scour Elevation for Check Flood

Abutment 1 163 163 163 162.5

Abutment 2 163 169 166.5 166

The micropiles were also designed to resist horizontal earth pressures resulting from scouring of the stream side only—this condition assumes complete failure of the existing retaining walls in service, strength, and extreme limit states. Though conservative, this condition is relevant assuming soil arching between pile locations occurs following a scour event, therefore leaving the micropiles acting similarly to a cantilever retaining wall. The estimated horizontal earth pressure due to 14 feet and 15.5 feet of soil is considered for the 100-year design flood at Abutment 1 and Abutment 2, respectively; 14.5 feet and 16 feet of soil is considered for the 200-year check flood, respectively. Active horizontal earth pressures were used for the scour condition due to anticipated larger deflections; at-rest horizontal earth pressures were used for the non-scour condition due to smaller anticipated deflections. A minimum differential hydrostatic pressure equal to 3.0 feet of water, applied on the retained side, was also considered in this condition as recommended in the AASHTO LRFD specifications.

4.5 Micropile Foundation Design

4.5.1 Geometry and Micropile Layout

Dimensions of the proposed pile cap at each stub abutment are presented in Table 4-5. In order to avoid installing micropiles through the existing abutment wall foundations, the proposed pile caps have a setback between 4 to 6 feet behind the existing abutment walls.

Table 4-5 – Proposed Pile Cap Dimensions

Abutment Bottom of Cap

Elevation (ft)

Width (ft)

Approximate Length1

(ft)

Depth (ft)

Abutment 1 169.7 5.0 40.4 4.0

Abutment 2 173.4 5.0 48.7 4.0 1 Pile cap length as used in analyses is approximated due to irregular geometry at the cap ends.

The proposed micropile layout consists of two rows of ten piles for each abutment as shown in Figure 4-2.

The center-to-center spacing in the y-direction (width of cap) is 31 inches for both abutments; the micropiles on the bridge side are expected to clear the existing abutment foundation by approximately 3 to 4 inches. The center-to-center spacing in the x-direction (length of cap) is 46 inches at Abutment 1 and 58 inches at Abutment 2. Due to the existing wraparound wall at the southwest end of Abutment 1 and known 1-in to 1-ft batter on the backside of the existing wall, an offset of approximately 52 inches from the face of the wall to center of the first pile is proposed; this should allow the piles to clear the existing wall, however, the piles will be installed through the existing footing.

Figure 4-2: Proposed pile layout at Abutment 1 (left) and Abutment 2 (right).

Additionally, there are site constraints imposed at the southeast end of Abutment 1 due to the existing upward grading between the Military Road NW Bridge abutment wall and the retaining wall along the

Luzon branch—a cheek wall extending from the new abutment to the existing abutment wall of Joyce Road is proposed to accommodate the grading. The cheek wall will be integrated with the pile cap and structurally designed to counter earth pressures of the slope. The earth pressures on the southeast end of the pile cap were also accounted for in the micropile design since the southwest end of the pile cap will extend to the face of the existing retaining wall, therefore having no soil for lateral resistance at the cap end.

4.5.2 Axial and Structural Analysis

The axial geotechnical resistance of the micropiles was evaluated following axial resistance analyses of a single micropile according to AASHTO LRFD specifications (2017) and the FHWA Micropile Design and Construction reference manual (2005). An axial group efficiency of 1.0 was used. Only the competent rock was assumed contributing to the axial resistance. The results indicate that, for the most critical case, the factored axial resistance of the micropile exceeds the anticipated total factored vertical load resulting from weight of the micropile, vertical load components of the superstructure and live loading, and weight of the pile cap. Based on the analyses, minimum 5 feet of rock bonded length is required to support the superstructure axially. The structural integrity of the micropile was also evaluated and met design requirements. Table 4-6 summarizes the results of the axial and structural resistance analyses. Appendix F presents the results and the detailed analyses.

Table 4-6 – Summary of Results for Axial and Structural Resistance Analyses

Abutment Abutment 1 Abutment 2

Results of Axial Analysis – Strength I

Estimated tip elevation 144 feet 135 feet

Minimum casing plunge length 1 3.5 feet 3 feet

Minimum bonded length 6.5 feet 7 feet

Factored geotechnical axial resistance at strength limit (per pile) 201 kips 160 kips

Maximum factored compression load on single micropile 187 kips 155 kips

CDR for geotechnical axial resistance 1.08 1.03

Results of Structural Analysis – Strength I

Capacity-to-demand ratio (CDR) for buckling 5.5 6.1

CDR for factored punching shear of pile cap – compression/tension 1.9/4.7 2.4/8.3

CDR for factored shear resistance 11.0 14.4

CDR for combined axial and flexural resistance 1.4 1.4

CDR for combined axial and flexural resistance at casing joint 2 1.1 1.1

Notes:

1 Geotechnical axial resistance along casing plunge length is conservatively neglected.

2 A minimum length of 5 feet was assumed between threaded casing joints.

4.5.3 Lateral Analysis

Lateral resistance analyses were performed for both abutments considering the entire abutment components. As for the axial component, only the resistance from the competent rock was considered in the analyses. In addition to the lateral load from breaking, wind, and temperature loads, earth pressure from the retained soil behind the abutment and the back wall, and a permanent surcharge pressure resulting from traffic are considered.

FB-MultiPier, a non-linear finite element analysis program, was used to model the behavior of the foundation group subjected to combined axial and lateral loads (group analysis). The analyses were performed to simulate long-term condition after the completion of the bridge construction. The most critical drained soil strength parameters were used to derive horizontal earth pressure behind the abutment at final condition. An equivalent 2 feet of fill (250 psf) of traffic surcharge was considered.

The applicable service and strength limit state load combinations from AASHTO LRFD Bridge Design Specifications (2017) were used for analysis. In accordance with AASHTO LRFD specifications, the scour condition due to 100-year design flood was considered for service and strength calculations. Additionally, one extreme event case was analyzed assuming a full scour condition due to a 200-year check flood. For each scour case, the micropile foundation is expected to support horizontal earth pressures due to the soil arching on the non-scour side.

Models of the entire abutment with the micropiles are considered in the analyses to simulate the behavior of the pile cap and pile elements. The p-y methods by Reese et al. (1974) for sand and Liang et al. (2009) for rock were used in FB-MultiPier. P-multipliers were applied to the pile p-y curves to account for the effects of group interaction, resulting in reduced lateral resistance of the soil. P-multipliers were determined based on pile spacing and values presented in Table 10.7.2.4-1 of AASHTO LRFD specifications. The micropile spacing ratio and corresponding P-multipliers are presented in Table 4-7.

Due to spacing constraints in combination with anticipated larger lateral loads and deflections of the scour condition, a maximum spacing of 31 inches (or 2.9 pile diameters) was used between pile rows at both abutments, while maintaining minimum 9-inch clearance from cap edge to pile edge in accordance with AASHTO LRFD specifications. Empirical data for pile spacings less than three pile diameters is very limited, however extrapolated values for a spacing of two pile diameters (0.45 for row 1, 0.33 for row 2, and 0.25 for row 3 or higher) as presented in the WSDOT Geotechnical Design Manual (2013) were referenced.

Accordingly, P-multiplier values of 0.76 (row 1) and 0.39 (row 2) were linearly interpolated based on the proposed pile spacing.

Table 4-7 – Summary of P-Multipliers Used in Lateral Analyses

Abutment Direction1 Spacing/Diameter Row 1 Row 2 Row 3+

Abutment 1 Length (Xp) 4.3 0.93 0.69 0.56

Width (Yp) 2.9 0.76 0.39 -

Abutment 2 Length (Xp) 5.4 1.00 0.85 0.70

Width (Yp) 2.9 0.76 0.39 -

Notes:

1 Refer to pile layouts presented in Figure 4-2.

Table 4-8 summarizes the results of the lateral analyses, and Appendix F contains the output of the FB- MultiPier analyses. All micropile loads and deflections are within allowable range under normal ground conditions. However, micropiles were designed for maximum moment, shear, and axial loads imposed during the scour condition. Corresponding deflections for all service, strength, and extreme limit states are less than 2 inches under scour conditions, which assumes the complete failure of the existing retaining wall occurs, the micropiles will retain soil on the non-scour side due to soil arching between pile, and the bridge is still in service. Deflections of the two piles intersecting the existing retaining wall footing at the southwest corner of Abutment 1 were also estimated; maximum deflections are less than 1.2” at the top of the footing (elevation 165 feet) and decrease with increased depth. Since all micropiles are terminated in bedrock, the vertical movement of the micropiles is expected to be negligible. Additionally, maximum total elastic compression and elastic extension are less 0.1 inches at both abutments.

Table 4-8 – Summary of Results for Lateral Analyses

Component Abutment 1 Abutment 2

Casing 10.75-inch-diameter and 0.595-inch thick

Reinforcement #14 #14

Micropile cased length 21 feet 33 feet

Micropile uncased length 6.5 feet 7.0 feet

FB-MultiPier Results – Scour Considered1

Maximum lateral deflection at pile top—Service I 1.24 inches 1.55 inch

Maximum lateral deflection at pile top—Service I without Scour 0.23 inches 0.13 inches

Maximum biaxial shear force on a single pile—Strength I 27 kips 21 kips

Maximum biaxial moment on a single pile—Strength I 169 kip-feet 180 kip-feet

Maximum biaxial shear force on a single pile—Extreme Event II 15 kips 19 kips

Maximum biaxial moment on a single pile—Extreme Event II 75 kip-feet 93 kip-feet

Notes:

1 The maximum moment, shear, and axial loads were analyzed for estimated scour conditions and using load combinations based on AASHTO 2017 LRFD Specifications.

4.5.4 Global Stability Analysis

Global stability analysis was performed for both abutments using the cross section at the centerline of the bridge. This represents the maximum height of the rear and forward abutments wall under the bridge.

Global stability analyses were performed using Slide2 software, version 9.0 (Rocscience, Inc., 2020). Both circular and noncircular failure surfaces were analyzed using drained soil strength parameters; due to the granular nature of the soils, undrained strength conditions are not anticipated. Spencer’s (1967) limit equilibrium method was used for the stability assessment and computation of the factor of safety. The Spencer method was selected over other methods because it satisfies both moment and force equilibrium according to FHWA GEC 4. The resistance of the micropiles was conservatively neglected since the new stub abutments are behind the existing abutment walls. A rapid drawdown analysis was also performed using the 100-year design flood elevation of 173.2 feet.

The CDR values of all the models against global stability failure is greater than 1.0 (i.e., φ factor of safety), as shown in Table 4-9. Appendix F contains the detailed analyses.

Table 4-9 – Summary of Results for Global Stability Analyses

Abutment Wall Failure Mode Strength Condition Factor of Safety CDR

Abutment 1

Circular Drained

2.0 1.3

Block 2.0 1.3

Circular Rapid Drawdown for 100-year Flood

1.6 1.0

Block 1.5 1.0

Abutment 2

Circular Drained

1.9 1.3

Block 1.8 1.2

Circular Rapid Drawdown for 100-year Flood

1.6 1.1

Block 1.5 1.0

4.6 Foundation Recommendations

The foundation recommendations for the abutments are summarized in Table 4-10.

Table 4-10 – Design Recommendations for Micropile Foundations

Structure Abutment 1 (South) Abutment 2 (North)

Casing size 10.75-inch diameter and 0.595-inch-thick 1

Reinforcing bar #14 threaded bar (Grade 75)

Compressive strength of grout 5,000 psi

Cut-off elevation 171.4 feet 175.0 feet

Minimum design tip elevation 144.0 feet 135.0 feet

Minimum total length 27.4 feet 40.0 feet

Minimum bond length in rock 2 6.5 feet 7 feet

Minimum casing embedment in rock 3.5 feet 3 feet

Notes:

1 Steel casing must conform to ASTM A252, Grade 3 with minimum yield strength of 60 ksi.

2 Minimum bond length is determined using geotechnical bond strength estimated based on the grout-to-ground bond and neglects end bearing. Minimum bond length must be verified by a verification load test prior to micropile production.

4.7 Erosion Protection

Erosion protection consisting of 2.0 feet Class 3 riprap and 0.5 feet granular material…