A1 ATT 2B - Geotechnical Report Bridge Replacement RFI 5 (22R0008).pdf

HUME LAKE ROAD BRIDGE REPLACEMENT PROJECT

GEOTECHNICAL REPORT

Prepared for:

United States Department of Agriculture Forest Service

Prepared by:

MAY 1, 2019

1291S8-18-D-0009, TASK ORDER 1

MGE JOB NO. 585

TABLE OF CONTENTS

Page

1.0 INTRODUCTION

1.1 Purpose

1.2 Geotechnical Services

2.0 PROJECT LOCATION

3.0 SITE DESCRIPTION

3.1 South Approach

3.2 Existing Bridge Structure

3.3 North Approach

4.0 AS-BUILT FOUNDATION DATA

5.0 PROJECT DESCRIPTION

5.1 Bridge Structure

5.2 Bridge Approaches

6.0 FIELD EXPLORATION AND TESTING

6.1 Utility Locating

6.2 Borings

6.2.1 Sampling

6.2.2 Hammer Energy and Blow Counts

6.2.3 Logging and Surveying

6.3 Laboratory Testing

6.4 Seismic Refraction

7.0 SITE GEOLOGY AND SUBSURFACE CONDITIONS

7.1 Site Geology

7.2 Subsurface Conditions

7.2.1 Roadway and Bridge Abutment Fill

7.2.2 Recent Alluvium

7.2.3 Igneous Rock

7.3 Groundwater

8.0 SCOUR EVALUATION

9.0 CORROSION EVALUATION

10.0 SITE SEISMIC CONDITIONS

10.1 Ground Motion

10.2 Liquefaction Potential

10.3 Seismically Induced Settlement

10.4 Lateral Spreading

10.5 Elastic Settlement

10.6 Surface Fault Rupture

10.7 Seismic Slope Instability Bridge Approach Embankments

11.0 DISCUSSION AND CONCLUSIONS

TABLE OF CONTENTS (cont.)

12.0 DESIGNER’S FOUNDATION DATA AND LOADING

12.1 Cast-in-Drilled-Hole (CIDH) Foundations

13.0 RECOMMENDATIONS

13.1 Cast-in-Drilled-Hole (CIDH) Piles

13.1.1 Compressive Resistance

13.1.2 Tension

13.1.3 Foundation Design Recommendations and Pile Data Table

13.1.4 Lateral Resistance

13.1.5 Settlement

13.2 Lateral Earth Pressures

14.0 APPROACH FILL EARTHWORK

14.1 Earthwork

14.2 Overexcavation

14.3 Fill Material

14.4 Expansive Soil

14.5 Slope Geometry and Stability

14.6 Settlement

15.0 CONSTRUCTION CONSIDERATIONS

15.1 Site Preparation

15.2 Fill Placement and Compaction Requirements

15.2.1 Slopes

15.2.2 Benching

15.3 Existing Utilities

15.4 Cast-in-Drilled Hole (CIDH) Pile Foundations

15.5 Excavation and Shoring

15.6 Dewatering

16.0 PLAN REVIEW AND CONSTRUCTION OBSERVATION

17.0 LIMITATIONS

18.0 REFERENCES

TABLE OF CONTENTS (cont.)

TABLES

1 Pile Foundation Design Data 2 USGS and Caltrans Fault Data 3 USFS’s Foundation Design Data Sheet 4 Foundation Factored Design Loads 5 Foundation Design Recommendations 6 Pile Data Table 7 LPile Parameters 8 Equivalent Fluid Pressures

FIGURES

1 Vicinity Map 2 Site and Exploration Plan 3 Design ARS Curve 4 Liquefaction Analysis Summary – Boring B-1 5 Liquefaction Analysis Summary – Boring B-2 6 Estimated Axial Shaft Resistance, Abutment 1 7 Estimated Axial Shaft Resistance, Abutment 2

APPENDICES

A Subsurface Explorations B Geotechnical Laboratory Testing Procedures and Results C GeoVision’s Report of Geotechnical Investigation D Caltrans Standard Plan and Bridge Detail Sheets E Long Meadow Creek Bridge, As-Built Plans F Important Information About Your Geotechnical/Environmental Report

MGE Project No. 585 585 Hume Lake Road Bridge Replacement FR.docx

FOUNDATION REPORT

HUME LAKE ROAD BRIDGE REPLACEMENT

HUME LAKE RANGER DISTRICT

SEQUOIA NATIONAL FOREST, CALIFORNIA

1.0 INTRODUCTION

1.1 Purpose

MGE Engineering completed a limited study of foundation conditions at the above site in accordance with our Task Order 1 agreement dated November 21, 2018 under contract 1291S8-18-D-0009 between MGE Engineering (MGE) and the USDA Forest Service (USFS). The purpose of this study is to collect soil and rock samples, provide geotechnical laboratory testing services, provide soil classifications and general engineering characteristics of soil and rock, and provide geotechnical recommendations for the proposed bridge foundations and the earthwork associated with bridge construction. Pavement design and slope stability analyses are not part of our task order scope of services. The proposed bridge will replace the existing bridge.

1.2 Geotechnical Services

To prepare this report, MGE Engineering completed the following:

Discussed the project and consulted with Mr. Jon Sampson of USFS.

Reviewed published geologic maps and available literature pertaining to the site.

Reviewed three As-built plan sheets dated May 23, 1946 and titled, “Long Meadow Creek Bridge Hume Lake, Sequoia National Forest” and sheet names of “Location,” “Bridge Details,” and “Abutment Details.”

Completed a site review on December 10, 2018.

Performed borings on December 10 and 11, 2018.

Subcontracted GeoVision to perform seismic refraction and shear wave imaging on December 12 and 13, 2018.

Reviewed the GeoVision seismic survey report.

Reviewed two Preliminary Plan sheets, received electronically on March 4, 2019, dated March 4, 2019, and titled, “Hume Lake Road Bridge Replacement – Proposal,” and “Hume Lake Road Bridge Replacement – Proposal-Loads.”

Performed engineering and geologic evaluations and analyses.

This report supersedes all previous data, opinions, reports and recommendations provided by MGE. Limitations of this study are discussed in the section “Limitations.”

2.0 PROJECT LOCATION

The existing bridge is located on Hume Lake Road where it crosses over the Long Meadow Creek in the rural community of Hume, Fresno County, California. Site coordinates are approximately latitude 36.78713 N and longitude -118.913552 W. The bridge is centrally located within the Hume Lake Christian Camp (HLCC) and is one of three crossings over the creek in this area. MGE understands that during summer HLCC operations, over 2,000 people attend the camp each week and that the bridge serves as the main crossing for the camp (especially in the summer months).

MGE assumes that the project limits extend from approximately 50 feet north of the bridge to 50 feet south of the bridge. The site location is shown in Figure 1, Vicinity Map. Figure 2, Site and Exploration Plan, shows the existing bridge location and the locations of our two borings performed at the site.

3.0 SITE DESCRIPTION

At the bridge location, the creek flows easterly into Hume Lake within a trapezoidal channel. The trapezoidal channel is present from approximately 50 feet west of the bridge and extends 200 feet downstream to the lake. The channel is relatively flat and the channel bottom at the bridge is generally at elevation 5,193 feet. Riprap protection is present on the channel banks from the top of bank to the bank toe. Coarse sand, gravel, cobble, and boulder size material is present in the channel bottom.

Land use near the bridge site includes private residences and commercial structures, office buildings, and recreational facilities associated with the HLCC. A fueling station is north of and within 100 feet of the bridge. Underground utilities are present near and attached to the bridge and trees are near the northeast and southwest corners of the bridge. Areas beyond the top of the channel bank (except southeast of the bridge) are generally paved and provide parking areas for the HLCC.

3.1 South Approach

The south bridge approach has two lanes and a wye intersection with a service road that approaches from the west. South of the bridge the road is named 10 Mile Road. On the east side of the road, a swimming pool and associated pool maintenance building facilities are present and west of the road parking areas and HLCC buildings are present. 10 Mile Road turns east approximately 100 feet south of the bridge and parallels the creek alignment on the south side of the pool.

3.2 Existing Bridge Structure

The existing bridge is a single-span structure that is approximately 40 feet long by 26 feet wide.

The bridge has an asphalt paved concrete deck that has two 10-foot-wide lanes and 2-foot 11-inch wide curbs. The bridge has two 6-foot-wide pedestrian walkways on either side of the bridge structure that were added to the bridge by the HLCC. The substructure consists of reinforced concrete wall abutments. The existing bridge deck is generally flat across the span and matches the top of channel bank elevation. As-built plans show the deck elevation as 5547 feet and the lake high water elevation of 5542 feet (assumed NGVD 1929 datum).

3.3 North Approach

The existing approach consists of two lanes located between the parking lot for the HLCC administration building on the east and the fueling station and general store on the west. A driveway entrance to the administration parking area is present at the northeast bridge abutment corner. The north approach slopes gently towards the bridge with the end of the vertical curve near the north bridge abutment.

4.0 AS-BUILT FOUNDATION DATA

The USFS provided as-built bridge plans via email for our study. The existing bridge is shown founded on 20-foot long Douglas Fir timber piles that are 12-inches diameter. Two rows of 5 timber piles are shown at each abutment. Piles have transverse bridge direction center-to-center spacing of 4-feet, 4 inches and longitudinal bridge direction center-to-center spacing of 2-feet 6 inches. Plan notes on the “Abutment Detail” sheet indicate driving piles to 16.5 ton bearing.

Timber pile cut-off elevation is shown at 5539.9 feet and bottom of pile cap elevation at 5538.9 feet. The bridge stone masonry wingwalls are shown supported on 12-inch thick reinforced concrete spread footings.

The plans show that the bridge is designed for H-20 loading and is designed to the 1944 A.A.O.S.H.O specifications. The elevations on the plans are assumed to be referenced to NGVD 1929 datum.

5.0 PROJECT DESCRIPTION

The following project description is based on our understanding of the project, and information about the replacement structure and alignment provided by the USFS on February 6, 2019 and revised using preliminary plan sheets provided on March 4, 2019.

The project involves replacing the existing bridge structure on the same alignment.

The roadway grade will remain the same The bridge will be 54-feet, 10-inches long and 24-feet wide The bridge will have two 11-foot, 8-inch-wide lanes Retaining walls are not required Pavement design at the approaches will match existing Foundations will consist of 30-inch diameter cast-in-drilled-hole (CIDH) piles

5.1 Bridge Structure

The USFS indicates that the bridge will be a single-span reinforced concrete voided slab with the above dimensions. The substructure will consist of seat type abutments supported on four CIDH piles at each abutment. MGE names the proposed abutments from south to north as Abutment 1 and Abutment 2, respectively.

Table 1 presents details of the proposed bridge support locations and foundations.

Table 1 Pile Foundation Design Data

Support Number

Design Method

Diameter and Pile Type

Finish Grade

Elevation

(feet)

Bottom of Pile Cap Elevation

(feet)

Cut-off Elevation

(feet)

Pile Cap Size (feet)

Permissible Settlement

Under Service

Load (inch)

Number of Piles per Support B L

Abutment 1 (south)

LRFD

30-inch

CIDH

5,204.25 5,199.3 5,199.8 4.0 27.42 1.0 4

Abutment 2 (north)

LRFD

30-inch

CIDH

5,204.25 5,199.3 5,199.8 4.0 27.42 1.0 4

Notes:

Data provided by USFS on march 4, 2019.

CIDH = cast-in-drilled hole LRFD = Load and Resistance Factor Design

5.2 Bridge Approaches

the bridge approach fill will not increase the road grade profile and MGE anticipates that fill will be limited to the abutment wall backfill and roadway base used for the design pavement section.

MGE is not aware of any proposed retaining walls and we assume that retaining walls will not be required at the bridge approaches. MGE expects that the existing abutment fill and bridge structure will be removed as part of the improvements.

6.0 FIELD EXPLORATION AND TESTING

6.1 Utility Locating

USFS personnel used white paint to mark proposed boring areas on the pavement near the bridge abutments. The USFS called Underground Services Alert (USA) to report the proposed boring areas for utility location and received a USA ticket describing the proposed field activities and the utilities present at the site. MGE forwarded the USA ticket (received electronically from USFS’s Jonathan Berry on December 3, 2018) to Geo-Ex Subsurface Exploration (Geo-Ex) so that Geo-Ex could also notify USA of the locations for utility clearance.

6.2 Borings

We evaluated subsurface conditions for the proposed bridge foundations by drilling, sampling, and logging two borings designated as boring B-1 and boring B-2. We selected boring depths to collect data for both shallow and deep foundation options.

Geo-Ex drilled on December 10 and 11, 2018 and completed two borings for this study; one boring at either end of the bridge near the roadway centerline. Geo-Ex encountered refusal in each boring when using solid auger in Boring B-1 and when using a side-discharge fish-tail mud-rotary bit in Boring B-2. Below refusal, Geo-Ex used diamond core drilling tools to drill into and recover rock. Please refer to Figure 2, Site and Exploration Plan, to view a site plan of the bridge and the locations of the borings.

6.2.1 Sampling

MGE personnel recovered disturbed samples from the borings by means of a lined 1.4-inch inside diameter (I.D.) Standard Penetration Test (SPT) sampler advanced with standard 140-pound calibrated automatic hammer with a 30-inch drop (per ASTM D1586). MGE personnel used a lined 2.4-inch I.D. California sampler driven in a similar manner to the SPT sampler to recover disturbed soil samples. We lined the samplers with stainless-steel liner tubes. MGE did not take Shelby Tube samples due to the very loose and loose apparent density of the granular overburden soil.

MGE recovered the soil and rock to perform visual classification and evaluation, and to perform soil and rock laboratory tests. MGE’s field representative recorded percent rock core recovery and percent Rock Quality Designation (RQD) for the recovered rock cores. We transported soil and rock samples to our office for further classification and transported selected rock samples to Gulf Shore Construction Services for laboratory testing.

6.2.2 Hammer Energy and Blow Counts

Geo-Ex reported that the hammer energy ratio (ERi) for the automatic hammer used for sampling is ERi=77%. The hammer was last calibrated on October 14, 2015. Geo-Ex did not perform a hammer energy calibration for this project.

We present uncorrected blow counts (per ASTM D 1586) recorded in the field on the boring logs.

We used Burmister’s Energy-Area correction (Burmister, 1948) and the hammer energy reported by the driller to calculate N60 blow counts. We also show the N60 blow counts on the boring logs and used the corrected blow counts to determine the appropriate apparent density descriptions for the soil. Please refer to Appendix A, Subsurface Explorations, where we further describe our field procedures and present the boring logs.

6.2.3 Logging and Surveying

An MGE field representative logged the recovered soil and rock samples and soil cuttings. The field representative used characteristics such as consistency; color; gradation; sampler penetration resistance; observations of samples and cores; percent sample recovery, rock recovery and RQD;

and observation of auger/drill cuttings to classify and log the soil and rock. The field representative measured groundwater levels within the boring during drilling. Each test boring was backfilled with cement grout after completion.

MGE Engineering used an existing nail and shiner set in the asphalt on the south side of the bridge as a temporary benchmark and reference for the boring elevations and used an assumed elevation of 5,205 feet for the nail and shiner. We measured the boring locations from the existing bridge and estimated the latitude and longitude boring coordinates using an iPhone.

MGE corrected the iPhone boring coordinates in the office using Google Earth Pro and the known boring locations measured from the bridge. We present the boring locations on Figure 2, Site and Exploration Plan and the coordinates of the boring logs, descriptions of encountered materials, dates of exploration, and laboratory test locations in Appendix A, Subsurface Explorations.

6.3 Laboratory Testing

MGE coordinated with Mr. Stephen Romero (USFS) and Mr. Jon Sampson (USFS) via e-mail between December 19 and December 21, 2018 to develop the laboratory test schedule. The USFS elected to test the recovered rock core samples and not test the soil samples. Some of the laboratory tests indicated in the task order scope of services were not performed because of material types and sampling methods.

Tests eliminated from the geotechnical laboratory testing program include:

10 - Moisture Content - Unit Dry Weight (ASTM-D2216 – ASTM-D4767) 4 - Sieve Analysis (ASTM-D6913) 4 - Plasticity Index (ASTM-D4318) 1 - Remolded Direct Shear Test (ASTM D3080) 1 - Optional Consolidation with time rate measurements (ASTM D2435)

Gulf Shore Construction Services in Rancho Cordova, California, performed the following tests on selected rock cores.

• Unconfined Compressive Strength of Rock (ASTM D7012)

• Point Load Index (ASTM D5731)

Results of laboratory testing are included in Appendix B, Geotechnical Laboratory Testing Procedures and Results.

6.4 Seismic Refraction

Geovision performed geophysical surveys at the Hume Lake Road Bridge site on December 12 and 13, 2018. Geovision mobilized on December 11 and demobilized on December 14, 2018.

Geovision completed two seismic lines (Lines 1 and 2) using Rayleigh and Love wave multichannel analysis of surface waves (MASW), linear passive surface waves, and shear wave (S-wave) refraction. Both lines were 235 feet long with 48 geophones placed at 5 feet spacing and each aligned in a linear array. The noise source for the Rayleigh wave MASW was a truck-mounted accelerated weight drop (AWD) system and sledgehammers. The noise source for the S-wave refraction and Love wave MASW was a sledgehammer fitted with a “golf shoe” or traction plank.

Geovision completed Line 1 in the USFS approved location between the gift shop/market and the north side of the channel. Line 2 was offset from the south side of the channel to the asphalt road.

AstroTurf is present in the original Line 2 location that prevented use of that location. Any deployed noise sources would have damaged the AstroTurf surface and therefore, the line was moved away from that location.

Wooden/plastic decking adjacent to the volleyball court and the swimming pool is present in the alternate locations proposed for the southern line. The decking prohibited Geovision from deploying shots or geophones in that location, shortened the seismic line in that area and, therefore, did not present an adequate line length that is required to achieve the project objectives. The large grass area to the east of the wood deck and south of the lake (not pre-approved) was deemed too far from the bridge for testing. The AstroTurf and wooden/plastic decking were not discernable from aerial photos prior to arriving at the site.

Rock velocities ranged between 5,600 fps to greater than10,500 fps. Recent alluvium and fill velocities ranged between 1,100 fps to 2,300 fps. Typically, seismic refraction velocities greater than 7,500 fps may be difficult to excavate depending on the contractor’s equipment and means and methods. Rock rippability will also depend on the rock hardness, fracture spacing and orientation, degree of weathering, and other factors.

We present Geovision’s seismic refraction and seismic shear wave interpretations in Appendix C, Geovisions Report of Geophysical Investigation.

7.0 SITE GEOLOGY AND SUBSURFACE CONDITIONS

7.1 Site Geology

At the bridge site, a published geologic map (Moore and Nokleberg, 1992) shows surface geology as the Cretaceous Giant Forest Granodiorite. The map describes the rock as a medium-grained equigranular, hornblende-rich granodiorite with abundant mafic inclusions. Although present, the geologic map does not show Holocene Alluvium mapped in the creek channel.

Rock outcrop observed during our site geologic reconnaissance of the project area is generally consistent with the Giant Forest Granodiorite descriptions in the published mapping. In general, the rock outcrop is not present within the immediate bridge vicinity. However, observed outcrop within the HLCC appears massive, fresh, and unfractured. Unconsolidated sand, gravel, cobbles, and boulders are present within the active creek channel. MGE did not observe rock within the channel near the bridge.

We did not observe evidence of significant geologic hazards (such as remnant landslides, severe erosion, very soft soils, etc.) during this study.

7.2 Subsurface Conditions

The borings encountered fill and alluvial soil layers over rock. Rock encountered in the borings appears igneous and is generally consistent with surface geologic map descriptions. We interpret three units at the site: roadway fill, recent alluvium, and igneous rock. Refer to our boring logs in Appendix A, Subsurface Explorations, for more detailed descriptions of the soil and rock units.

7.2.1 Roadway and Bridge Abutment Fill

We encountered the roadway/bridge abutment fill from ground surface to depths of 7 to 10 feet. The fill generally consists of very loose and loose silty sand and clayey sand.

7.2.2 Recent Alluvium

Native soil was encountered between the fill and rock to depths of 28 to 38 feet in the borings.

The native material is generally very loose to medium dense poorly-graded sand and poorly-graded sand with silt.

GeoVision’s seismic refraction surveys included in Appendix C shows refraction velocities indicative of alluvial deposits to 25 to 30 feet below ground surface. In our opinion, this unit is susceptible to scour/erosion where exposed and is not suitable for supporting bridge foundations.

7.2.3 Igneous Rock

We encountered igneous rock within each boring. The igneous rock is below the recent alluvium.

The rock encountered in the borings is generally massive, decomposed to moderately weathered, is moderately fractured, and has local very intensely fractured zones. Fractures generally dip at 50 and 60 degrees. Zones of decomposed rock are present where the rock is very intensely fractured.

Laboratory tests indicate correlated uniaxial compressive strengths range between 0.23 to 26 kips per square inch (ksi) based on Point Load Index tests and between 0.9 to 12 ksi based on uniaxial compressive strength tests. In our opinion, the rock mass is generally scour-resistant and can provide foundation support for the proposed bridge structure.

7.3 Groundwater

During our subsurface exploration (December 2018), we encountered ground water at a depth of about 9 to 10 feet in the borings. In each boring, the encountered groundwater had a similar elevation compared to the surface water level in the creek channel. Appendix A, Subsurface Explorations, presents tables showing depth and elevation of groundwater we observed in the borings, and surface water elevations we measured during our fieldwork.

We note that the lake surface elevation shown on topographic mapping (USGS, 1992) is 5,201 feet. The topographic map lake surface elevation is approximately 4 feet below existing roadway grade at the bridge location. Historic photographs in Google Earth imagery show backwater conditions in the channel when the lake surface elevation is up.

Groundwater and surface water levels can fluctuate due to changes in precipitation, tributary creek levels, irrigation, pumping, drought, dam releases, and other factors.

8.0 SCOUR EVALUATION

The USFS indicates a total scour depth to 1-foot – elevation 5,194.00 feet – at the bridge abutments. Long term and short-term scour estimates, that include degradation, contraction scour, and local scour, were not provided.

In our opinion, soil within the channel and encountered in the borings is not scour-resistant. We incorporated the provided scour depth into our pile capacity calculations for abutment support.

9.0 CORROSION EVALUATION

The USFS Task Order did not include laboratory testing to evaluate corrosion potential. MGE did not perform corrosion tests to evaluate the fill and alluvium. MGE retained soil samples collected from the borings and we can perform corrosion tests through an outside lab, if the USFS desires.

10.0 SITE SEISMIC CONDITIONS

10.1 Ground Motion

MGE Engineering used the USGS US Design Maps (USGS, 2018) online web-based tool to obtain seismic data for the site. The USGS tool uses the 2009 American Association of State Highway and Transportation Officials (AASHTO) Guide Specifications for LRFD Seismic Bridge design to calculate probabilistic acceleration response spectra. The nearest deterministic fault that contributes to the probabilistic seismic hazard is shown in Table 2.

Table 2 USGS and Caltrans Fault Data

Fault Parameters Kern Canyon Fault

USGS Fault Identification Number (FID) 494

Maximum Moment Magnitude (Mmax) 7.51

Fault Type Normal1

Fault Dip (degrees) 601

Dip Direction East

Site-to-Fault Distance, RRUP 48.5 kilometers / 30.1 miles Note:

1 MGE obtained these fault characteristics from Caltrans ARS Online.

GeoVision estimated that the average shear-wave velocity for the upper 100 feet of the subsurface profile (VS30) is about 1,203 feet per second (fps) and 1,423 fps using their seismic data from Line 1 and Line 2, respectively. MGE averaged the two results and rounded down to obtain a Vs30 of 1,300 fps for this site. MGE used the average value to develop the ARS curve for the site. A VS30 = 1,300 fps corresponds to a Soil Profile Type C per AASHTO LRFD Bridge Design Specifications, Table 3.10.3.1-1 – Site Class Definitions.

We used AASTHO Tables 3.10.3.2-1, 3.10.3.2-2, and 3.10.3.2-3 and the USGS seismic data to select site factors Fpga, Fa, and Fv, respectively. We then used the seismic parameters obtained from the USGS and AASHTO Equations 3.10.4.2-1 through -6 to calculate the elastic seismic coefficient (Csm) for structure periods from 0.01 to 5 seconds.

Based on the above information, we recommend the structure design use the following parameters:

Shear Wave Velocity, VS30: 1,300 fps Soil Profile Type C Deterministic Fault Contributing to Probabilistic Hazard: Kern Canyon Fault Maximum Moment Magnitude (Mmax) of Contributing Hazard: 7.5 Peak Ground Acceleration (PGA): 0.21 g Controlling Spectra: USGS Probabilistic Spectrum

Figure 3, Design ARS Curve, presents the recommended design AASHTO ARS Curve.

10.2 Liquefaction Potential

Liquefaction is a secondary effect associated with seismic loading. It can occur when saturated, loose to medium dense, granular soils or low plasticity cohesive soils are subjected to ground shaking.

However, loose to medium dense, granular soil layers in the alluvium unit will likely be below groundwater based on backwater conditions in the creek when the lake surface elevation is high and the encountered groundwater levels during this study. The poorly graded sand with silt and poorly graded sand layers below the proposed approach fill have a high potential for liquefaction. We would expect liquefaction to occur between elevation 5,201 to 5,178 feet based on the borings. We include Figure 4, Liquefaction Analysis Summary, Boring B-1 and Figure 5, Liquefaction Analysis Summary, Boring B-2, that present our liquefaction analysis based on data collected from borings B-1 and B-2.

Although the potential for liquefaction at the site is considered high, in our opinion, the potential for liquefaction to affect the bridge foundation axial capacity at the site is low because the foundations will be founded in rock. Affects to lateral pile capacity from liquefaction is addressed in Section 10.4, Lateral Spreading.

10.3 Seismically Induced Settlement

During a seismic event, ground shaking can cause densification of granular soil that can result in settlement of the ground surface. The potential for seismically induced settlement affecting the bridge foundation axial capacity at this site is considered low due to foundations bearing in rock.

However, we estimate that approximately 15 to 18 inches of seismically induced settlement could occur at the boring locations based on the soil profile encountered in the borings. This amount of settlement during a seismic event could affect the bridge approaches and create downdrag loads on the pile foundations. Estimated downdrag loads are approximately 35 to 50 kips negative nominal side resistance between elevation 5,198 feet to 5,178 feet. We present recommendations to reduce the amount of seismically induced settlement in the earthwork section of this report.

10.4 Lateral Spreading

Lateral spreading is caused by cyclic loading of liquefied soil and the incremental displacements that accumulate during loading. Typically, lateral displacements move towards a free face like a bridge embankment slope or creek bank. Lateral spreading displacements can range from a few inches to tens of feet depending on the characteristics of the ground motion, liquefied soil, and other factors. Lateral spreading can include phenomena such as flow liquefaction. Flow liquefaction can occur when a slope becomes unstable under static loading due to strength loss caused by liquefaction. Surface expressions of flow liquefaction generally include soil slumping and large deformation.

MGE evaluated lateral displacement using boring data and our soil classifications. Our evaluations used different approaches to provide an estimated lateral displacement. Estimated displacements range between 2 to 6 feet. We include LPile parameters in Table 6 Pile Data Table, that the USFS can use to model the lateral loads due to the liquefied soil condition.

However, in our opinion, the potential for lateral spreading to significantly affect the bridge foundations at this site is considered low due to:

the liquefiable material is generally located below the channel bottom, the restraining effects of the pile foundations and the abutment wall, the removal and replacement of the bridge abutment fill per our following recommendations, and the anticipated low level of ground motion at the site.

10.5 Elastic Settlement

We expect only nominal elastic settlement will occur for approach roadway and abutment backfill if existing materials are removed and replaced with engineered Structural Backfill as we recommend.

10.6 Surface Fault Rupture

No active fault traces are shown on published mapping to cross the site, and the site is not within or adjacent to an Alquist–Priolo Earthquake Fault Zone for fault-rupture hazard. Therefore, we conclude the potential for surface fault rupture within the project limits is low.

10.7 Seismic Slope Instability Bridge Approach Embankments

In our opinion, the potential for seismic slope instability of the existing creek banks to affect the bridge structure is considered low. We expect that the approaches will have a similar configuration to existing, and that fill slopes will be stable if constructed at 1V:2H slopes and our earthwork recommendations are followed during construction.

11.0 DISCUSSION AND CONCLUSIONS

In our opinion, the rock at the site should have adequate bearing support for the proposed bridge foundations. We recommend founding the bridge on deep foundations that bear in rock. In our opinion, the subsurface conditions appear suitable for constructing the proposed 30-inch-diameter CIDH piles at the abutments.

We anticipate groundwater will be encountered during construction at the abutments. At each support, surface water diversion, diking, and/or cofferdam construction may be required, depending on creek levels in the channel. Slurry drilling methods likely will be required to install the CIDH piles. The alluvium includes very loose and loose sand, which will require temporary casing within the upper 10 to 15 feet below the pile cap (possibly deeper depending on actual encountered conditions and pile cut-off elevation) for CIDH installation, ground control, and fluid containment outside the channel. We present our geotechnical design recommendations for 30-inch-diameter CIDH piles in subsequent sections.

Loose alluvial materials present at the existing abutments have the potential to liquefy and settle during a seismic event. These materials will be below the proposed roadway approaches and could cause ground distortion during a seismic event. The loose materials will require removal and replacement to reduce potential settlement.

12.0 DESIGNER’S FOUNDATION DATA AND LOADING

12.1 Cast-in-Drilled-Hole (CIDH) Foundations

Foundation data and loading for 30-inch-diameter CIDH piles provided by the USFS are presented in Table 3 andTABLE 4. The USFS provided Service Limit loads only. This information is the basis for our pile calculations and recommendations.

Table 3 USFS’s Foundation Design Data Sheet

Support Number

Design Method

Diameter and

Pile Type

Finish Grade Elevation

(feet)

Cut-off Elevation

(feet)

Pile Cap Size (feet) Permissible Settlement

Under Service

Load (inch)

Number of Piles per Support B L

Abutment 1 LRFD 30-inch-

CIDH

5,204.25 5,199.8 4.0 27.42 1.0 4

Abutment 2 LRFD 30-inch-

CIDH

5,204.25 5,199.8 4.0 27.42 1.0 4

Notes:

Data provided by the USFS on March 4, 2019.

Table 4 Foundation Factored Design Loads

Support Number

Service-I Limit State (kips) Strength Limit State

(Controlling Group, kips) Extreme Limit State

(Controlling Group, kips)

Total Load Permanent

Loads Compression Tension Compression Tension

Per Support

Max.

Per Pile

Per Support

Per Support

Max.

Per Pile

Per Support

Max.

Per Pile

Per Support

Max.

Per Pile

Per Support

Max.

Per Pile

Abut-1 273 NP 137 NP NP 0 0 NP NP NP NP

Abut 2 273 NP 137 NP NP 0 0 NP NP NP NP

Notes:

Max. = maximum NP = Information not provided

13.0 RECOMMENDATIONS

13.1 Cast-in-Drilled-Hole (CIDH) Piles

13.1.1 Compressive Resistance

MGE Engineering calculated the axial compressive resistance for 30-inch-diameter CIDH piles using the AASHTO LRFD Bridge Design Specifications (AASHTO, 2017) and the Federal Highway Administration (FHWA) GEC 010 – Drilled Shafts: Construction Procedures and LRFD Design Methods (FHWA, 2010). Our analyses neglected end bearing contributions for the CIDH piles. We calculated bearing capacity using only side friction and discounted materials within the upper 5 feet of the pile and above the estimated scour elevation of 5,194 feet.

For the CIDH piles, a geotechnical resistance factor (φ) of 0.5 was used to evaluate compressive resistance at the Strength Limit State consistent with the LRFD method. A geotechnical resistance factor (φ) of 1.0 was used to evaluate compressive resistance at the Extreme Limit

State.

We included negative skin friction at the abutments from seismically induced settlement in our axial bearing calculations. We present a graphical representation of axial capacity versus depth in Figure 6, Estimated Axial Shaft Resistance, Abutment 1 and Figure 7, Estimated Axial Shaft Resistance, Abutment 2.

13.1.2 Tension

The USFS did not provide any tension demands for the bridge foundations.

13.1.3 Foundation Design Recommendations and Pile Data Table

The foundation recommendations for the abutment piles are summarized in Table 5, and the recommended pile data for inclusion in the project plans is presented in Table 6. MGE understands that the USFS will provide the lateral pile-tip elevations and include those within their plan set. If the USFS determines that lateral pile tip elevations are lower than the compression tip elevation shown in Table 5 and Table 6, the USFS should use the lateral tip elevation as the specified tip elevation for pile installations.

Table 5 Foundation Design Recommendations

Su pp or t L oc at io n

D ia m et er a nd

Pi le

T yp e

C ut

-o ff

E le va tio n (f ee t)

Service-I Limit State Load Per Support

(kips)

To ta l P er m is si bl e Se ttl em en t ( in ch es

Required Factored Nominal Resistance3, 4

(kips)

D es ig n

Ti p

El ev at io ns

(f

Sp ec if ie d Ti p El ev at io ns

(f

Strength Extreme

Total Permanent

Comp5 ϕ = 0.5

Tens ϕ = 0.4

Comp ϕ = 1.0

Tens ϕ = 1.0

Abut 1 30”

CIDH 5,199.8 273 137 1.0 NP 0 N/A N/A

5,165(a) N/A (b) 5,169 (c) TBD (d)

5,165

Abut 2 30”

CIDH 5,199.8 273 137 1.0 NP 0 N/A N/A

5,165(a) N/A (b) 5,169 (c) TBD (d)

5,165

Notes:

1 Design tip elevations are controlled by (a) Compression, (b) Tension, (c) Settlement, (d) Lateral.

2 The specified tip elevations shall not be raised without consultation with the geotechnical engineer.

3 Required nominal resistance in compression is per pile.

4 Loads are rounded up 5 Strength Limit State Compression load is estimated based on four CIDH piles per abutment support and rounded up to the nearest 10 kips.

N/A = not applicable NP = Information Not Provided Tens = Tension Comp = Compression Abut = Abutment TBD = To Be Determined by the USFS

Table 6 Pile Data Table

Support Diameter and

Pile Type

Nominal Resistance (kips)

Design Tip Elevations1, 3

(feet)

Specified Tip Elevations2

(feet) Compression Tension

Abutment 1 30-inch CIDH 340 0

5,165(a) N/A (b) 5,169 (c) TBD (d)

5,165

Abutment 2 30-inch CIDH 340 0

5,165(a) N/A (b) 5,169 (c) TBD (d)

5,165

Notes:

1 Design tip elevations are controlled by (a) Compression, (b) Tension, (c) Settlement, (d) Lateral.

2 The CIDH specified tip elevation shall not be raised.

3 The USFS shall calculate and provide the lateral tip elevations.

13.1.4 Lateral Resistance

We provide soil and rock parameters for the USFS to evaluate lateral pile loading using LPILE software by Ensoft. We understand that the USFS will use this data to evaluate CIDH pile lengths and lateral loads for the abutments.

MGE Engineering performed drilled borings near the proposed abutments to evaluate soil and rock characteristics and rock surface elevations. GeoVision performed seismic refraction surveys and shear wave profiles to evaluate soil and rock characteristics. We used GeoVision’s interpreted seismic data to supplement our boring data in our evaluations.

The following Table 7, LPILE Parameters, presents soil and rock parameters for the USFS to use for lateral pile analyses. The parameters shown in the table represent 30 feet of loose alluvium over weathered and fractured rock.

Table 7 LPile Parameters

Elevation (feet)

Depth (feet)

LPILE

Soil Type (p-y curve model)

Total/Effective Unit Weight

(pcf)

Friction Angle (deg.)

Blow Count k Below/ Above Water (pci) qu (ksi)

Rock Mass

Modulus (ksi)

RQD

krm (dim.)

5,205 to 5,198

0 to

Medium Sand

(Reese) 120/58 32 20/25 -- -- -- --

5,198 to 5,178

15 to

Liquefied Sand

(Rollins) 120/58 5

5,178 to 5,155

27 to

Weak Rock

140/78 -- -- 0.75 500 25 0.0005

Notes:

1. Top of boring elevations are 5,205 feet.

2. Rock modeled as weak, fractured, and weathered igneous rock.

3. Compressive strength and rock mass modulus estimated from laboratory tests and observations. RQD percent is a weighted average for the borings.

4. To calculate buoyant soil weight, take the total unit weight and subtract 62.5 pcf.

dim. = dimension psf = pounds per square foot pcf = pounds per cubic foot ksi = kips per square inch % = percent ft = feet deg. = degrees qu = unconfined compressive strength RQD = rock-quality designation

13.1.5 Settlement

We estimate that the proposed 30-inch CIDH pile should settle less than ½-inch under the proposed service loads. The settlement should occur essentially as the loads are applied. We do not anticipate significant long-term settlement at this site. MGE performed settlement analysis for a single pile and not a pile group.

13.2 Lateral Earth Pressures

We recommend using the equivalent fluid pressures (EFPs) shown in Table 8 to design the abutment and wingwalls.

Table 8 Equivalent Fluid Pressures

Active Static EFP (pcf)

At-Rest EFP (pcf)

Passive Static EFP (pcf)

Active Seismic EFP (pcf)

34 58 900 10 Notes:

EFP = equivalent fluid pressure pcf = pounds per cubic foot

For static design, apply the resultant of the static active earth pressure at a depth of H/3 from the bottom of the wall where H equals the wall height in feet. For seismic design, use a uniform distributed pressure increase using the active seismic EFP.

The EFP values shown above assume level backfill conditions using Structural Backfill with a soil unit weight of 120 pcf, a minimum angle of internal friction of 34 degrees, and drainage behind the abutments is placed in accordance with either Caltrans Bridge Details BO-3 or B9-06 (Caltrans, 2018a). We include these bridge detail sheets in Appendix D, Caltrans Standard Plan and Bridge Detail Sheets. Structural Backfill should meet specifications described in Section

704.04 of the Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects FP-14 (FP-14), (US DOT, 2014).

Apply the greater of the following surcharge loads:

0.3 times the calculated surcharge pressure, or A minimum surcharge pressure of 240 psf.

14.0 APPROACH FILL EARTHWORK

14.1 Earthwork

Site grading and earthwork should be performed in accordance with Sections 200, 250, 300, 400 and 700 of FP-14.

14.2 Overexcavation

We recommend removing and replacing the existing fill and native material from the top of the pile cap to ground surface and to a minimum 10-feet behind the abutment backwall and wingwalls. MGE should review the excavation to confirm the materials and make recommendations for additional overexcavation, as necessary. The purpose for the overexcavation, replacement, and recompaction is to provide a firm foundation for the approach fill, and to reduce the potential for liquefaction, elastic settlement, and seismically induced settlement at the approaches.

14.3 Fill Material

Imported fill should be approved by the geotechnical engineer and should conform to Structural Backfill criteria in Section 704.04 of FP-14 for abutment backfill. The Contractor should provide laboratory test results that show the imported material:

Meets the gradation requirements of the specification, and Has a Plasticity Index less than 6.

Imported fill should be free of debris and organic material.

14.4 Expansive Soil

Do not use expansive soil as fill. Expansive soil is a soil with an Expansion Index greater than or equal to 50.

14.5 Slope Geometry and Stability

In our opinion, the near-surface soils can provide adequate support for proposed approach fill heights assuming the Contractor follows FP-14 Standards and Specifications and our recommendations for subgrade preparation herein. We expect new embankments with exterior slopes at 1V:2H or flatter will be stable if constructed per FP-14 Specifications and per our recommendations.

14.6 Settlement

The alluvial soil should experience nominal elastic settlement due to the proposed fill thickness behind the abutments. We anticipate most settlement should occur as the fill is placed.

15.0 CONSTRUCTION CONSIDERATIONS

This section is provided for the benefit of specification writers, construction personnel, and contractors. This section identifies pertinent sections of the AASTHO LRFD Bridge Design Specifications, FP-14 Standard Specifications, Cal/OSHA, and other design criteria that we assumed in developing our geotechnical design recommendations. This section provides information to identify subsurface conditions that may be encountered in the field during construction. For the project described herein, we recommend that this report, the report appendices, and subsequent addenda be included with project documents during the bidding process for reference purposes.

The applicability of our recommended design parameters is contingent on good construction practice. Poor construction techniques may alter conditions from those upon which our recommendations are based, and therefore result in poor performance. Our analyses assumed that this project is constructed according to FP-14 U.S. Customary Units (FHWA, 2014) construction Standard Specifications. The following sections provide additional construction considerations for this project.

15.1 Site Preparation

Prior to site grading, surface vegetation and surficial topsoil containing organics should be removed. The depth of this removal is expected to be approximately 3 to 12 inches. Trees or shrubs to be removed should include the entire root ball and all roots larger than ½-inch-diameter. This may require laborers handpicking the roots from the subsurface soils prior to compaction. Deleterious materials and vegetation should be disposed offsite.

In addition, construction areas should be cleared to a depth necessary to remove all surface and subsurface structures associated with current development of the site, including underground utilities and any other deleterious materials. Topsoil and organic rich soils can be re-used to cover side slopes and landscaped areas once the approach fill is constructed with imported granular materials.

To prepare the site for construction, fill, deleterious debris, vegetation, and other surface and subsurface items associated with development at the site should be removed from areas that will be excavated, receive fill, or be constructed upon. Following site clearing, the subgrade in areas designated to remain at grade or receive fill should be scarified to a depth of at least 6 inches, moisture-conditioned to within two percent of optimum moisture, thoroughly mixed, and uniformly compacted to the appropriate level of compaction.

We expect that the Contractor can excavate the soil and rock to the indicated depths using typical heavy-duty construction equipment. If hard rock is encountered at a shallow depth, MGE should be immediately contacted so that MGE can evaluate the rock and make supplemental recommendations for subgrade and site preparation.

MGE’s review of the prepared excavations is a condition of our recommendations. Our review will allow for supplemental recommendations as needed, such as adding drainage, removing unsuitable material, etc., as necessary.

15.2 Fill Placement and Compaction Requirements

All fill and backfill should be placed and compacted to a dense/firm and unyielding condition as specified in accordance with FP-14 Sections 209.09 and 209.10. Fill and backfill should be moisture conditioned to within 2 percent of the optimum moisture content and compacted to at least 95 percent of the Modified Proctor maximum dry density within the wall backfill area and within 30 inches of finished grade below pavement areas.

Fill and backfill should be placed in maximum 6-inch-thick lifts when using heavy compaction equipment. The lift thickness should be reduced to 4 inches when using small compaction equipment for backfilling against walls.

If subgrade or placed engineered fill soil becomes loose or disturbed during grading, the affected soil should be excavated to expose competent, undisturbed soil and replaced with properly compacted engineered fill. We recommend that MGE personnel observe engineered fill placement and perform in-place density tests to evaluate whether the specified compaction is achieved.

If time passes due to construction delays, or otherwise, and the site is left exposed, the Contractor, upon restarting construction, should recondition and recompact the soil to depths necessary to meet the project specifications.

15.2.1 Slopes

We understand that slopes less than 5-feet high will be constructed as part of the project. We recommend 1V:2H or flatter slopes to satisfy global stability requirements. We recommend that fill slopes be seeded and/or planted with relatively fast-growing vegetation to reduce the potential for surface sloughing and erosion.

15.2.2 Benching

We recommend new fill is benched onto existing fill slopes or natural slopes exceeding 1V:3H in accordance with FP-14 Section 204.

15.3 Existing Utilities

Any existing utilities should be moved or protected in place prior to construction of pile foundations and bridge structure elements to avoid any construction conflicts.

15.4 Cast-in-Drilled Hole (CIDH) Pile Foundations

The subsurface conditions encountered in the borings include groundwater above expected specified pile tip elevations. Therefore, we anticipate a contractor would install CIDH piles by the “wet” method (FP-14, Section 565.06.2b), including slurry drilling and concrete placed under slurry. The borings encountered granular layers (sand and gravel) within CIDH pile foundation depths. The Contractor should review the boring logs (Appendix A) and plan accordingly.

Temporary casing may be required to elevation 5,178 feet during construction of the CIDH piles to help mitigate caving soil and to contain drill fluid during the excavation process. Casing may be required below elevation 5,178 feet depending on actual rock elevations and the Contractor should be prepared to use additional casing, as necessary, to maintain the CIDH excavations.

Consideration for CIDH pile installation/excavation for this site/project includes the following:

Caving soils due to presence of loose granular layers within pile foundation depths, Presence of groundwater within pile foundation depths, and

Requirement for temporary casing during pile excavation/construction for ground control and to facilitate slurry drilling in the channel.

The contractor should construct CIDH piles in conformance with Section 565 – Drilled Shafts, of the FP-14 Standard Specifications. For CIDH piles with center-to-center spacing less than 3.0 diameters, the sequence of shaft installation should be specified in the contract documents. Also, do not drill holes or drive casing for an adjacent pile until 20 hours have elapsed after concrete placement in the preceding pile; do verify that prequalification test results for the concrete mix design show that the concrete will attain at least 1,800 pounds per square inch compressive strength at the time of drilling.

Permanent casing should not be used for CIDH installation because the piles would not meet the required nominal bearing resistance due to reduced skin friction. The Contractor is responsible for the design of temporary casing, including actual length(s) and diameter(s), to install CIDH piles according to the above specifications without defects or damage to existing utilities/ facilities. The temporary casing must be non-corrugated with smooth surfaces and the diameter should be at least 8 inches greater than the CIDH pile to help prevent binding of the drilling tool. Installing the temporary casing below the specified pile tip elevation is not permissible.

15.5 Excavation and Shoring

Site safety is the responsibility of the Contractor. During construction, temporary construction backslopes should be reviewed by a competent person to evaluate stability and to possibly require supplemental support (e.g., local shoring in areas of soft/weak materials).