Geo_Rpt_15-19_Lake_Crescent_Road_Final_Geotechnical_Report_30Dec2015.pdf

Rehabilitate Route 11, Lake Crescent Road (US Highway 101)

WA NPS OLYM 11(9)

Geotechnical Report No. 19-15

Rehabilitate Route 11, Lake Crescent Road (US Highway 101)

Final Geotechnical Recommendations

Olympic National Park

Clallam County, Washington

December 2015

Geotechnical Report GR19-15 Page i

Contents

1. INTRODUCTION

1.1 Project Location

1.2 Scope of Project

1.3 Background Information

2. PHYSICAL SETTING

2.1 General Site Characterization and Topography

2.2 Lake Surface Levels, and Average Temperatures and Precipitation

2.3 Geology

2.4 Seismicity

3. GEOTECHNICAL INVESTIGATIONS

3.1 Subsurface Investigation

3.1.1 Drill Equipment, In-Situ Testing, and Descriptive Conventions

3.1.2 Subgrade Test Borings

3.1.3 Geotechncial Test Borings

3.2 Laboratory Testing

3.3 Test Boring Instrumentation

4. GENERAL SITE CONDITIONS AND GEOTECHNICAL RECOMMENDATIONS

4.1 General Observations

4.2 Surfacing Section and Subgrade

4.3 Subgrade Stabilization

4.4 Unstable Embankment Slope Sites

4.5 Retaining Wall Discussion and Recommendations

4.5.1 Liquefaction

4.5.2 Station 110+10 Fairholm Store Wall

4.5.3 Station 203+90 Retaining Wall

4.5.4 Station 406+00 and 409+30 Retaining Walls

4.5.5 MSE Wall Design and Recommendations

4.6 Riprap Revetment

4.7 Ditches

5. CONSTRUCTION ADVISORY FOR SUBSURFACE STABILIZATION AND

REINFORCEMENT MITIGATION AREAS

5.1 Subsurface Stabilization and Reinforcement Mitigation Limits

5.2 Subsurface Conditions and Temporary Excavations

5.3 Dewatering

6. LIMITATIONS

7. REFERENCES

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Appendices

APPENDIX A

Figures

APPENDIX B

Tables

APPENDIX C

Unified Soil Classification System Soil Descriptive Terminology Rock Descriptive Terminology Subgrade Test Boring Logs Geotechnical Test Boring Logs

APPENDIX D

Rock Core Photography

APPENDIX E

Laboratory Test Reports

APPENDIX F

Project Pictures

APPENDIX G

Historical Pictures

Geotechnical Report GR19-15 Page iii

Figure 1: Project Vicinity Map.

Figure 2: Project Location Map.

BOP

EOP

Project Vicinity

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1. INTRODUCTION

The proposed Rehabilitate Route 11 Lake Crescent Road Project will preserve and extend the service life of the highway and provide safety improvements to approximately 12.3 miles of US Highway 101 in Olympic National Park. US Highway 101, located along Lake Crescent, is a key transportation corridor, providing scenic and recreational access as well as providing direct connection to Port Angeles, WA. (Puget Sound) and Forks, WA. (Northern Washington State Coastline). (Figure 1)

1.1 Project Location

The beginning of the project begins near Olympic National Park’s west boundary, approximately 28 miles west of Port Angeles, WA and approximately 1.7 miles west of Camp David Jr. Road.

The project extends east approximately 12 mile along Highway 101 to the Park’s east boundary.

(Figure 2)

1.2 Scope of Project

The project includes rehabilitation of the existing pavement throughout the length of the project, subgrade and shoulder stabilization, replacement of existing retaining walls at specific locations, and fill slope erosion mitigation.

1.3 Background Information

The Olympic Highway was constructed in the 1920’s along the south shoreline of Lake Crescent.

The original construction method is interpreted to be partial bench to sidecast, to full benching and typical of methods used at the time. Where sidecast techniques were used, it is believed that the materials were sidecast onto the slope with little regard for proper benching, compaction, and thorough removal of organics. It has been interpreted that the resulting sidecast fill remained in a loose, unconsolidated state, and in a condition that promoted downhill creep over time. Historic pictures show that timber crib retaining walls were constructed during the initial construction of the road. Historic pictures are provided in Appendix G.

It is likely that the roadway was widened and straightened and the timber crib walls were replaced with concrete cribs walls sometime between the initial roadway construction and the early 1980’s. In the 1980’s WFLHD contracted a highway project through the same area, which included spot improvements to the horizontal alignment, improvement to cut slopes, and construction of gabion basket retaining walls.

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2. PHYSICAL SETTING

2.1 General Site Characterization and Topography

The project begins at Station 33+00 (BOP), just east of Sol Duc Hot Springs Road at an approximate elevation of 1130 feet. The road descends close to Lake Crescent surface elevation of about 580 feet. The road has numerous curves between Stations 145 and 420 and between Station 480 and end of the project (EOP) at station 646+27. The road also exhibits a somewhat rolling terrain profile between Stations 225 and 552, and then begins to climb to an elevation of 780 feet at the EOP. The majority of the project follows Lake Crescent’s southern shoreline. Olympic National Park, Barnes Point Visitor Center, is located at Barnes Point, near Station 465. Private single occupancy homes are located west of the Visitor Center between Stations 420 and 435.

Steep cut slopes with many stream crossings characterize the general topography of the project. Douglas fir and cedar trees are numerous and large. The current top of pavement width is about 24 feet, including shoulders. There are several areas with a passing lane and numerous small pullouts.

2.2 Lake Surface Levels, and Average Temperatures and Precipitation

Olympic National Park provided daily average lake level data recorded from October 2003 to September 2015. This data shows over a twelve-year period the lake level cyclically varied about 3.5 feet.

The Olympic Peninsula has a moderate marine climate with summers that tend to be fair and warm; July and August are normally the driest months. Winters are mild at lower elevation, with temperatures in the 30s and 40s; snowfall can be heavy in the mountains, with accumulations of up to 10 feet. A Western Regional Climate Center (WRCC) monitoring station is located at the Elwha Ranger Station (COOP# 452548) located about 10 miles east of the end of the project at an elevation of 380 feet. A second WRCC monitoring station is Forks 1E (COOP# 452914) located about 23 miles west-southwest of the beginning of the project. The average maximum and minimum temperatures, precipitation, and snowfall recorded at each station are summarized in Tables 1 and 2. Some data is rounded to the nearest tenth. The Elwha Ranger Station weather station is on the east side of the Olympic Mountains and typically receives less precipitation than the Forks 1E weather station.

Table 1: Average maximum and minimum temperatures and monthly precipitation at Elwha Ranger Station, Washington, COOP# 452548 between 07/1/1942 and 12/31/2014.

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec Annual Max. Temp. (°F) 41.2 44.4 50.2 56.5 62.4 66.9 72.5 73.8 67.6 56.2 44.5 40.7 57 Min. Temp. (°F) 32.4 32.9 34.8 37.8 42.6 46.9 50.3 51.2 47.5 41.4 35.1 32.4 40 Precip. (in.) 9.3 6.9 6.2 3.6 2.2 1.5 0.9 1.0 1.7 5.6 10.6 10.2 56 Snowfall (in.) 6.5 1.8 1.1 0 0 0 0 0 0 0 1.2 3.4 14

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Table 2: Period of Record Monthly Climate Summary for WRCC Forks 1E, Washington, COOP# 452914 between 11/1/1907 and 01/19/2015.

Jan Feb Mar Apr May Jun Jul Aug Sep Oct Nov Dec Annual

Max. Temp.(°F) 44.9 48.7 51.6 56.9 62.9 66.9 71.6 72.5 69.2 60.3 50.6 45.5 59 Min. Temp. (°F) 33.4 34.2 35.3 37.5 42 46.4 49.3 49.6 46.7 42.3 37.3 34.3 41 Precip. (in.) 17.8 13.6 13 8.6 5.3 3.5 2.3 2.5 4.8 11.7 16.4 18.6 118 Snowfall (in.) 5.6 2.1 2.1 0.2 0 0 0 0 0 0 0.7 2.3 13

2.3 Geology

The Olympic Mountains are a mostly Tertiary-aged subduction complex comprising underplated deep-marine siliciclastic rocks and the overlying marine basalt and autochthonous marginal marine basin fill. Between these two provinces lies the Puget Lowland, which was invaded repeatedly in Pleistocene time by ice of the Cordilleran ice sheet that covered the contact between the North Cascades and Olympic Mountains provinces with unconsolidated sediments as much as 3280 feet thick. The Cordilleran ice sheet extended to where Lake Crescent is now located. Lake Crescent was inundated by ice during the last maximum glacial ice extent that occurred approximately 25,000–21,000 years ago during the Late Wisconsin-age. (Schasse, 2003) Figure 3 shows an excerpt of the Geologic Map of the Washington Portion of Port Angeles 1:100,000 Quadrangle (Schasse, 2003) where the project is located.

Based on the geologic map the surficial geology between the BOP and about station 110 consists of Pleistocene-age Vashon Recessional Outwash (Qgo) of the Fraser Glaciation.

Generally, between stations 110 and about 400 the roadway is underlain with varying thicknesses of fill that overlies Eocene and Paleocene-age marine sedimentary rocks of the Blue Mountain Unit (ERm). This rock unit contains gray to black lithic sandstone, siltstone, argillite, granule or pebble conglomerate, and siltstone or slate-clast breccia; commonly laminated and rhythmically bedded; locally contains very thick sandstone beds.

Between about station 400 to nearly the EOP the geologic map shows middle and lower Eocene-age marine basaltic rocks of the lower Crescent Formation (Evcp) underlie the roadway and fill material. This rock unit is dominated by tholeiitic pillow basalt, but also contains basalt breccia, and volcanoclastic sandstone and conglomerate.

The Barnes Point area contains surficial deposits of Holocene-age alluvium (Qa) consisting of well-sorted cobble and pebble gravel, sandy gravel gravelly sand, silt, and clay. An area near Sledgehammer point is mapped as Holocene and Pleistocene-age landslide deposits (Qls) that contain clay, silt, sand, gravel, and larger blocks deposited by mass wasting which may be unstratified, and chaotic.

Near the end of the project bedrock consists of middle Eocene-age Aldwell Formation (EM2a) consisting of thin to medium-bedded, well-indurated marine siltstone and sandy siltstone with

Geotechnical Report GR19-15 Page 5 December 2015 interbeds of fine sandstone. Massive lenses of pebbles, cobbles and boulders of basalt occur sporadically, as well as lenses of basalt breccia and pillow lava.

Figure 3: An excerpt from Geologic Map of Washington Portion of Port Angeles 1:100,000 Quadrangle (Schasse, 2003) where the project is located.

Holocene-Age Alluvium (Non-glacial)

Holocene and Pleistocene-age Landslide Deposits (Non-glacial)

Pleistocene-age Vashon Recessional Outwash (Fraser Glaciation)

Eocene and Paleocene-age marine sedimentary Rock (Blue Mountain Unit)

Middle and lower Eocene-age marine basaltic rocks, pillow dominated (Crescent Formation)

Middle Eocene-age marine siltstone and sandy siltstone (Aldwell Formation)

Maximum extent of the Cordilleran ice sheet; barbs point towards ice front

Thrust fault- sawteeth on upper plate; dashed where inferred, dotted where concealed.

2.4 Seismicity

Northwest Washington State contains numerous active faults within 50 miles of the project site. A series of thrust faults are mapped along the north edge of the Olympic Peninsula. The Cascadia Subduction Zone is located approximately 100 miles west of the project, which according to the USGS has the potential to produce up to a magnitude 9.0 seismic event.

BOP

EOP

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3. GEOTECHNICAL INVESTIGATIONS

The Western Federal Lands Highway Division (WFLHD) conducted a geotechnical desktop study consisting of review of previous WFLHD highway project plan sets covering the project limits, published geologic maps, Washington State Department of Transportation (WSDOT) State Route Web videography, and interviews with the WSDOT Maintenance Supervisors for this section of Highway 101. In addition, several field reconnaissance trips took place between March 2014 and October 2015, and two subsurface investigation programs were conducted.

3.1 Subsurface Investigation

WSDOT was contracted by WFLHD to conduct two drilling investigations. Drilling investigations were conducted in July 2014 and May 2015.

Thirty subgrade test borings were drilled along the project alignment on Highway 101 between Olympic National Park boundaries. Soil samples and in-situ testing data from the subgrade test borings was used to characterize the existing pavement structural section. In general, subgrade test boring locations were spaced at ½-mile intervals and drilled near the center of the lane, generally alternating between the right and left lanes with the exception of a couple of the borings.

Thirteen geotechnical test borings were drilled in May 2015 to collect subsurface information for the purpose of characterizing soil, rock, and subsurface groundwater conditions. Six slope inclinometers and seven groundwater measuring instruments were installed at that time to monitor the subsurface condition over time.

Locations of the subgrade test borings and the geotechnical test borings are shown in Figures A- 1 through A-21, Appendix A, and are also listed in Table B-1 and Table B-2, Appendix B.

3.1.1 Drill Equipment, In-Situ Testing, and Descriptive Conventions

The 2014 subgrade test borings were drilled using a CME 55 truck-mounted drill (WSDOT 9C1-3) and a CME 850 track-mounted drill (WSDOT 9C2-3). The 2015 geotechnical test borings were drilled using a CME 55 track-mounted drill (WSDOT 9C7-1) and a CME 45 portable drill (WSDOT 9C4-2). All of the drills are equipped with calibrated auto hammers for Standard Penetration Tests (SPT).

Subgrade test borings were drilled using a 14 to 16-inch outside diameter (O.D.) flat profile earth bit attached to a 5-foot solid stem auger with an O.D. of 8 or 10 inches depending on the capability of the drill. The test borings extended to depths ranging from 4 feet to 7 feet below the road surface. Bulk samples were collected from the auger cuttings and boring sidewalls. A 6-inch diameter asphalt core was obtained using an asphalt core bit through the full depth of asphalt.

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The geotechnical test borings were advanced using a 5-foot hollow stem auger with an O.D. of 6 or 8-inches depending on the capability of the drill. If the auger met refusal prior to reaching the intended termination depth, an HQ3 coring system was used to advance the borings to termination depth. Geotechnical test borings range in depth between 11 feet and 71.4 feet.

Split-spoon drive samples were obtained in the subgrade and geotechnical test borings at select intervals using a standard 2-inch split spoon sampler (2-inch O.D.) or a 2.5-inch O.D. split spoon sampler to obtain disturbed soil samples and determine relative density and consistency of the soils. The SPT uses a 140-pound-hammer with a 30-inch drop to drive the split-spoon sampler 18 or 24 inches below the bottom of the borehole and requires recording the number of blows for each 6-inch increment of penetration. The bottom two blow counts for an 18-inch drive sampler and the middle two blow counts for the 24-inch drive sampler are added together to determine the blows per foot (bpf), which is commonly referred to as the field N-value.

A WSDOT drill inspector field classified and logged soil strata thicknesses and soil types in accordance to the Unified Soil Classification System (USCS). The samples were shipped to WFLHD in Vancouver, Washington, where the WFLHD Geotechnical Engineer reviewed the samples and the WSDOT draft subgrade test boring and geotechnical test boring logs.

Following review, selected samples of soil and rock were submitted to the WFLHD Materials Testing Laboratory for determination of index and physical properties. WSDOT draft boring logs were modified when necessary to reflect additional information obtained from laboratory testing or refinements to geologic interpretations or descriptions. The USCS visual-manual procedure (ASTM D2488) was used to characterize the soil units encountered during the drilling operations. Soil descriptive terminology, the USCS chart, and Rock Descriptive Terms are provided at the beginning of Appendix C.

Continuous rock cores were recovered, when possible, when the geotechnical test borings were advanced through bedrock or boulders using wet rotary methods with an HW casing advancer and HQ-sized casing in a triple tube system. Recovered rock core samples were described by the WSDOT drill inspector, placed into core boxes, photographed, and transported to the WFLHD Materials Testing Laboratory for further evaluation. Rock core was also photographed at the WFL Materials Laboratory and is presented in Appendix D.

3.1.2 Subgrade Test Borings

Subgrade test borings have a prefix designation of “SG.” For example, SG14-01 represents subgrade test boring (01) drilled in 2014 for this project.

Subgrade sampling procedures are to auger through the pavement without disturbing the underlying base rock. Once the asphalt and millings are swept away, the driller loosens the base rock with the auger without mixing the underlying materials. The field inspector hand excavates and places all the base rock into bulk sample bag(s). After hand cleaning the hole so no base rock is present, the driller carefully loosens the underlying material with the auger, Geotechnical Report GR19-15 Page 7 December 2015 avoiding mixing of individual, distinct layers. The field inspector then collects moisture samples from each distinct layer of differing material in addition to collecting additional bulk samples to a minimum depth of 5 feet or to auger refusal. Typically, two bulk samples from each layer are collected for subsequent testing.

Subgrade test boring logs are included in Appendix C in the order of project stationing progression from west to east.

3.1.3 Geotechncial Test Borings

“B” identifies geotechnical test borings. For example, B15-01 is geotechnical test boring (01) drilled in 2015 for this project. Geotechnical test boring logs are included in Appendix C in the order of increasing project stationing from west to east. Rock core photography is provided in Appendix D in the same order as the geotechnical test borings.

3.2 Laboratory Testing

This geotechnical report provides laboratory test results for selected soil and rock samples for the subgrade and geotechnical test borings taken to characterize the encountered materials.

Results of these laboratory tests are presented in Appendix E. A summary of the test results are shown in Tables B-1 and B-2.

Selected soil samples were tested at the WFLHD Materials Testing Laboratory in Vancouver, Washington in accordance with American Association of State Highway and Transportation Officials (AASHTO) and American Society of Testing and Materials (ASTM) testing methods.

These tests are used to obtain index properties, which allowed for soil classification and correlation between the tests and other physical properties of the soils as well as determination of soil support properties for pavement design. The following is a list of the index tests that were conducted and the AASHTO protocols that were followed:

• Sieve Analyses - AASHTO T 11/T 27

• Particle Size Analysis to 0.005 mm - AASHTO T 88,

• Atterberg Limits - AASHTO T 89/90

• Specific Gravity of Soils - AASHTO T 100,

• Natural Moisture Content - AASHTO T 265 (Sample dried at 230° F), and

• Mass Loss by Ignition – AASHTO T 267

In addition to the USCS, AASHTO soil classification system is also used within this report and in the laboratory test results. A description of the USCS and AASHTO soil classification systems are provided in Appendix C.

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3.3 Test Boring Instrumentation

Two types of geotechnical-monitoring instrumentation were installed in five of the geotechnical test borings. Slope inclinometer (SI) installations to monitor highway/slope movement and grouted-in vibrating wire piezometers (VWPs) with automated data loggers to monitor variations of groundwater levels. Geotechnical test boring logs have suffix designations of “si” or “vw” to identify SI and VWP installations, respectively. An open standpipe piezometer with a VWP is installed in B15-01, which is designated with the suffix “p” to identify the open stand pipe with vibrating wire piezometer.

The WSDOT Geotechnical Field Instrumentation Unit installed and initialized the data loggers and conducting quarter-annual site visits to perform measurements and download data. Data was sent electronically to WFLHD for evaluation, compiling, and storage of the data sets. Table B-2 provides information for the SI and VWP installations.

A SI installation consists of (2.75-inch) closed end, hollow pipe casing that is grouted within the annulus space of the test boring from the bottom of the hole to the ground surface where a locking monitoring well monument is installed to protect the borehole instruments. The casing contains four machined grooves at 90-degree angles that run the length of the casing on the inside. One set of grooves is typically oriented in line with the slope and the other set is perpendicular to the slope. After the grout has cured, a measuring probe (with wheel tracks) is periodically placed in the grooves and measurements are collected every two feet from the bottom of the casing to the surface. Periodic reading sets are compared to the first reading set (baseline reading set) to determine relative movement and depth of movement below the surface. The baseline reading set was taken on June 10, 2014. This initial data set is the base line for subsequent data sets. Subsequent data sets were measured on July 28 and November 19, 2015.

VWPs monitor groundwater levels by measuring pore-water pressure below the ground surface. After the grout has cured, an automated data logger is wired to the VWP cable to allow automated, periodic measurements to be taken. The data logger is housed in the locked well monument and can be accessed and connected on-site with a laptop computer to download the groundwater data.

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4. GENERAL SITE CONDITIONS AND GEOTECHNICAL RECOMMENDATIONS

This section discusses general site conditions based on site observations and test boring information and provides recommendations for subgrade stabilization, slope stability, and retaining wall design.

4.1 General Observations

There are numerous pavement patches throughout the project. Many of the patches appear to be a single event while other locations exhibit routine patching. WSDOT Area 3 maintenance supervisors Lyric Winn and Tim Roening stated that they often repair a pothole and find large voids (1 to 4 feet in diameter) extending down vertically or laterally under the roadway.

The outside (lakeside) lane appears to have settled in many areas. The settlement occurs along straight segments and at inside curves. (Pictures F-5, F-10, F- 14, F-15) The following are some possible reasons for roadway subsidence at various locations.

• The timber crib walls decomposed over time and the area was filled in with borrow.

• The concrete crib walls failed but were not replaced. Remnants of concrete crib logs can be found in the lake at various areas along the project.

• Ditch or groundwater is migrating under the roadway fill resulting in soil piping and loss of pavement structure support.

• Lake erosion undercutting the embankment slope resulting in loss of lateral support.

Some sections of roadway exhibit pavement distress in the form of random pavement cracking, rutting in wheel paths, slumped pavement, or some combination thereof. This type of distress is often indicative of poor subgrade conditions. However, based on the subgrade borings, the subgrade consists generally of granular material.

Many of the areas that exhibit pavement distress are in the left lane (outboard lane) and the outside shoulder is soft. It is interpreted that shore erosion and ground water piping under the road is the general cause of subgrade failure.

In other areas, pavement distress takes the form of tension cracking that is predominantly arcuate or sinuous longitudinally along the road. This type of distress is usually indicative of slope creep and/or settlement of fill materials.

4.2 Surfacing Section and Subgrade

Site observations with subsequent subgrade test boring and laboratory testing results indicate the following conditions under the roadway surface through the project.

The subgrade borings showed that the asphalt concrete thicknesses ranged from 5.5 inches to 21 inches. The base materials did not vary much from the underlying fill materials in most of

Geotechnical Report GR19-15 Page 10 December 2015 the subgrade borings. The materials underlying the pavement are generally granular soils that consist mostly of well-graded gravel with sand (GW), poorly-graded gravel with silty clay and sand (GP-GC), clayey gravel with sand (GC), well-graded sand with gravel (SW), silty sand with gravel (SM), poorly-graded sand with silt and gravel (SP-SM), and clayey sand with gravel (SC).

Silty gravel with sand (GM), gravel with silt and sand (GP-GM) and (GW-GM), well-graded sand with silt and gravel (SW-SM); poorly-graded SAND with clay and gravel (SP-SC) was also encountered deeper in some subgrade borings. One exception is SG14-01, which encountered asphalt grindings to termination depth of 5 feet.

Generally, the underlying pavement material was shown to have relative density ranging from medium dense to very dense; however, in the holes where a second SPT was performed deeper, the material exhibited lower relative density. Refusal was encountered in five of the subgrade holes, indicating the presence of possible nested cobbles or boulders. SG14-01 encountered a 1-foot deep void approximately 2 feet below ground surface. A void, 1.5 to 2 feet in diameter, was noted under the left edge of pavement that extended under the outside wheel path. Groundwater was not encountered in any of the subgrade holes during drilling or following withdrawal of the auger.

4.3 Subgrade Stabilization

The WSDOT Area 3 maintenance supervisors also stated that in the center of the road near Camp David Jr. Road (between approximately stations 97+50 and 100+50) water pumps through cracks in the pavement and their crews patch the pavement often. The affected area is under the left deceleration, intersection, acceleration turn lanes (in the direction of the project). Based on WSDOT’s concern in this area, the recommendation is to subexcavate 12 inches of the subgrade material, under the pavement structure, and replace with two layers of geogrid and select borrow. The stabilization geogrid is required to have an ultimate tensile strength of at least 1800 lb./ft. Refer to the final contract documents for the final construction requirements and typical details for subgrade reinforcement on the project.

We recommend constructing subgrade stabilization deep patches to mitigate the areas that exhibit subsidence and pavement distress. The approximate locations of the recommended subgrade stabilization deep patches are shown in Table B-3 in Appendix B. Subgrade reinforcement deep patches will be constructed either half-width or full-width and backfilled with select borrow. Stabilization geogrid is required to have an ultimate tensile strength of at least 1800 lb./ft. Refer to the final contract documents for the final construction requirements and typical details for subgrade reinforcement on the project.

4.4 Unstable Embankment Slope Sites

Three areas where road embankment instability is prevalent are between stations: 308+00 and 313+00, 589+00 and 597+00, and 633+00 and 635+25. These three areas are located on a relatively high, over-steepened embankment fill that exhibits a higher degree of patching, Geotechnical Report GR19-15 Page 11 December 2015 arcuate and sinuous longitudinal pavement cracks, and a wavy profile. This type of distress and patching is usually indicative of slope creep and/or settlement of fill materials. Although other areas in the project exhibit similar distress characteristics, these three areas were identified as exhibiting a higher degree of failure and perceived continuing road maintenance work.

In addition to the subsurface soil samples collected, SIs and VWPs were installed to provide additional information to identify the mechanisms that are occurring and characterize the resulting failures in these areas. Typically, a monitoring period of at least one year, depending on the rate of movement, provides adequate information for interpretation of subsurface movement. At the time of this report, the data currently collected is over a period of about five and half months with only one data set recorded after one significant rain event. Correlations between precipitation data and groundwater levels are inconclusive. SI data graphs are interpreted to show slight changes; however, additional data sets are required before a clear interpretation of the mechanisms can be made, and depths and rates at which failure may be occurring. Additional data reading sets are scheduled next year. Presentation of the monitoring data and interpretations will be discussed under a separate cover.

At the time of this report, a deep patch mitigation is recommended in the areas that are being monitored for slope displacements with the exception of between Stations 308+00 and 313+00.

This mitigation is similar to the subgrade stabilization deep patch design except they are 45 inches deep. This mitigation will not stop or reduce the rate of global slope displacement but rather is intended to reduce the rate at which pavement cracking and subsidence occur.

The approximate locations of the recommended deep patch mitigation are shown in Table B-3 in Appendix B. The deep patch is backfilled with select borrow and a stabilization geogrid having an ultimate tensile strength of at least 1800 lb./ft. is required. Refer to the final contract documents for the final construction requirements and typical details for subgrade reinforcement on the project.

4.5 Retaining Wall Discussion and Recommendations

Mechanically stabilized earth (MSE) retaining walls are proposed at the locations listed in Table B-4 in Appendix B. The site conditions and the subsurface conditions encountered at the wall locations are described in detail below. Three wall foundations will be at or below lake elevation. The foundation of the wall at Fairholm will be above the lake surface; however, the foundation will be subexcavated 3 feet below the bottom of the wall and backfilled with select granular backfill.

4.5.1 Liquefaction

Liquefaction occurs during a strong earthquake event when shaking causes the soil to rapidly lose shear strength due to excess porewater pressure, which results in the soil behaving like quicksand. Liquefaction of the soil can cause pipelines and underground utilities to rupture, foundations to move, and can pull roadways apart. Recently deposited and relatively loose

Geotechnical Report GR19-15 Page 12 December 2015 natural soils and un-compacted or poorly compacted fill soils are susceptible to liquefaction.

Submerged loose sand and silty sands are particularly susceptible to liquefaction. Loose silts and gravels may also have a potential for liquefaction. Dense, natural, well-compacted fills and cohesive soils are known to have a relatively low susceptibility to liquefaction.

When considering seismic hazards it is important to consider the tolerable level of risk relative to the perceived hazard. The risk associated with the seismic hazard is that during a large earthquake the wall suffers damage due to settlement and partial or complete replacement may be required. Another project risk that must be considered is the additional costs that would be associated with constructing a MSE wall with a foundation below the lake surface elevation, or deep foundation alternatives and/or ground improvement.

Liquefaction analyses were performed in accordance with AASHTO LRFD Specification, 7th

Edition (2014) using the software program Liquefy version 5.8 developed by CivilTech Software, Inc. Results from a seismic induced settlement analysis of the site where MSE walls are proposed are discussed within the individual wall subsection.

4.5.2 Station 110+10 Fairholm Store Wall

This site is located at the Fairholm Store parking area, at the western edge of Lake Crescent.

Two concrete cantilever or concrete counterfort retaining walls are supporting the pullout / store parking area, which extends close to the edge of Lake Crescent. The existing wall is shown in pictures F-1 through F-4 in the appendix. The two walls support the north and northeast side of the pullout/parking area with an interior angle of about 135 degrees. The northern facing wall is about 36 feet long and the northeast facing wall is about 40 feet long and the maximum heights of the walls are 12 feet at the junction. The two walls are founded on concrete footings that are about 8-10 inches thick with a toe that extends about 18 inches from the wall faces, as measured at the exposed footing at the junction of the two walls. The northeast facing footing is undermined about 1-foot beneath the footing caused by lake erosion, extending up to 2.5 feet below the bottom of the footing and from the junction extending 10 feet towards the road. The footing of the northeast wall appears to step up at least two times. The northern facing footing is buried under boulder size rocks except for about 3 feet at the junction. The northern facing wall inclination is about 3 to 4 degrees outward and the northeast facing wall inclination is close to vertical. There is a gap at the junction between the two walls of about 1.5 to 2 inches. The northern facing wall appears to be bowed-out when viewed along the top and outside face towards the northeast direction. The pavement in the pullout / parking area exhibits an uneven surface and surface water collects in the pavement distortions. The fill slope along the outside of the northeast wall to approximately station 111+00 is over-steepened due to lake erosion as well as from surface water runoff from the road into the lake.

Failure of one or both walls could result from further erosion undercutting the footings or from a seismic event causing the walls to topple over. The wall(s) may partially fail due to concrete degradation. There is a perceived risk to the integrity of the wall(s) if a vibratory compactor is used in the vicinity of these two current walls. Depending on the degree of failure of the wall, it

Geotechnical Report GR19-15 Page 13 December 2015 is not anticipated that both lanes of Highway 101 will be completely affected; however, damage to or failure of the left (west bound) lane is possible especially between stations 110+40 and 111+00. Failure of either wall will result in the loss of most of the pullout/parking lot.

Boring B15-01p encountered silty sand with gravel and silty gravel with sand, loose from 3.5 feet to 23 feet, then medium dense from 23 feet to 28 feet, and dense to very dense from 28 feet to

40.1 feet. The surface of the lake is approximately 18 feet below the boring collar.

The borings encountered an approximately 10-foot thick layer of loose to medium dense, saturated, gravel and sand below the bottom of the proposed MSE wall. This layer could be potentially liquefiable during a design-level earthquake with estimated settlements on the order of 4 inches. However, we are recommending a 4-foot deep subexcavation below the bottom of the MSE wall in order to construct the wall backfill on a firm foundation. This subexcavation should reduce the settlement to less than 2 inches, which should be tolerable for relatively flexible structures, such as MSE walls.

It should be noted that if the remaining medium dense saturated sand layer remains, the potential for horizontal displacements towards the lake due to lateral spreading may still exist.

Mitigations for reducing the lateral displacements could include full subexcavation and replacement, ground improvements (such as stone columns), or other structural solutions that would likely be cost prohibitive for this structure. Since this MSE wall is not directly supporting the roadway and should be relatively easy to repair by regrading, we have assumed that no mitigation is acceptable from a risk management standpoint. If liquefaction or lateral spread mitigation is required by the owner, we should be contacted for additional consultation.

If the wall is added to this project, we recommend the construction of a MSE wall to replace the concrete retaining wall. The wall foundation should be constructed 1-foot below the existing wall footing, however, based on the subsurface information obtained it is also recommended to subexcavate 4 feet below the bottom of the MSE foundation extending 3 feet in front and 5 feet behind the toe of the wall. This subexcavation will extend from the intersection of the two walls to one half of each wall length and will be encased with filter geotextile Class 1 Type A. A riprap revetment is recommended in front of the subexcavation.

A 45-inch deep patch was originally proposed under the left lane, if the MSE wall was not constructed, to provide some structural stability to the roadway in the event of complete failure of the wall. A deep patch from the end of the wall to Station 111+00 was also proposed due to the lake undercutting the embankment slope and resulting in an over-steepened embankment slope to reconstruct the shoulder for the guardrail. However, due to a concern of the vibration from the compaction operation affecting the stability of the existing wall the previous recommendations are retracted. The deep patch between the end of the wall and station 111+00 should be constructed if the existing wall at Fairholm is replaced.

Geotechnical Report GR19-15 Page 14 December 2015

4.5.3 Station 203+90 Retaining Wall

The existing concrete crib wall is about 10 feet high from the lake surface and the face is about 6 feet from the left edge of road. The bottom of the lake becomes very steeply inclined only 3 to 4 feet from the existing toe of wall. Viewing from the lake, the toe of the existing wall appears to be bearing on strong basaltic boulders, which do not appear compromised, but there is riprap placed in front of the toe that conceals some of the bearing area at the toe. Wall backfill is migrating through the wall face and the concrete cribbing is showing signs of degradation. The existing wall is shown in pictures F-7 and F-8.

Boring B15-02 was drilled through the back of the wall facing and encountered gravel. The boring was terminated at a depth of 11 feet due to drilling issues. Boring B15-03 was drilled in the right lane and encountered very dense sand with gravel between the depths of 1 to 3.5 feet. Very loose silty gravel was encountered between the depths of 3.5 feet and 8 feet, which overlies very dense gravel with sand to 10.5 feet. Slightly weathered marine sedimentary rock was encountered from 10.5 feet to termination depth of 31.5 feet.

A gabion faced MSE wall is recommended to replace the existing concrete crib wall at this location. This wall is expected to be founded on bedrock and cobble and boulder fill.

4.5.4 Station 406+00 and 409+30 Retaining Walls

Lake erosion affects are undermining the roadway embankment about 6 feet below the road at these two locations, extending under the embankment almost to the roadway edge. The existing embankment is about 7 to 8 feet high. Longitudinal pavement cracking has been observed between stations 409+50 and 410+50. Pictures F-17 through F-19 show this area.

Boring B15-08 was drilled close to the left fog line and encountered medium dense sand and very loose gravel with sand to a depth of 8.5 feet. Dense clayey sand with gravel was encountered to a depth of 28 feet, which is underlain by very dense sandy silt to test boring termination depth of 31.5 feet. Boring 15-09 was also drilled close to the left fog line and encountered loose to very loose silty gravel to a depth of 7.5 feet. Dense clayey sand was encountered below 7.5 feet and extending to the test boring termination depth of 26 feet. We estimate that the amount of total and differential settlement at this site would be less than 1-inch.

A riprap revetment was considered at these two locations; however, during a field review with the WFLHD Design Team it was determined that the slope leading to the lake bottom is too steeply inclined. For that reason and due to shore erosion a gabion faced MSE wall is recommended in order to provide support for the roadway.

4.5.5 MSE Wall Design and Recommendations

Conventional spread footings, MSE foundation system and driven piles are the foundation types that were considered for these four walls.

Geotechnical Report GR19-15 Page 15

A conventional concrete spread footing foundation was considered but not chosen because the foundation of a MSE wall is more flexible, is faster to construct, and is a more cost-effective choice. Driven piles were considered for the Fairholme wall site; however, due to the cost of construction, a deep foundation design was not chosen. A deep foundation design was not considered for the two walls between station 406+00 and 411+60 because the subsurface soils encountered in the borings are dense to very dense.

Gabion rock faced geogrid reinforced walls were designed for the project. The geogrid reinforcement is spaced 1.5 feet vertically. Design details will be provided directly to the Highway Designer.

The MSE walls were designed in accordance with AASHTO LRFD Specification, 7th

Edition (2014) using the software program MSEW 3.0 developed by ADAMA Engineering, Inc. Global stability analyses were performed using Slide 6.0 developed by Rocscience, Inc.

The soil parameters listed in Table 3 were used to design the MSE walls.

Table 3: Soil Parameters used for MSE Wall Design.

Location

Maximum Height

(ft.)

Reinforced Fill Retained Backfill Foundation Soil Unit

Weight (lb./ft.3)

Internal Angle of Friction (degree)

Unit Weight (lb./ft.3)

Internal Angle of Friction (degree)

Unit Weight (lb./ft.3)

Internal Angle of Friction (degree)

110+10 to 110+75* 14 125 34 125 34 115 31 203+90 to 204+40 10 125 34 125 34 130 36 406+00 to 407+40 8 125 34 125 34 110 28 409+30 to 411+60 8 125 34 125 34 110 28

*Optional wall Note: No cohesion assumed.

Reinforcement geogrid will be used in the MSE walls. The required ultimate tensile strength and long-term design tensile strength required for the geogrid reinforcement will follow the requirements outlined in Standard Specifications for Construction of Roads and Bridges on Federal Highway Projects (FP-14), 714.04 for Type III reinforcement geogrid.

The MSE reinforced zone and the area between the end of the reinforcement and the face of the temporary excavation should be backfilled with select granular backfill.

4.6 Riprap Revetment

WFLHD Hydraulics Team should be consulted for a riprap scour protection design at the toe of the walls.

Geotechnical Report GR19-15 Page 16 December 2015

4.7 Ditches

Generally between stations 109 and 420, and between station 480 and EOP the ditch system is not operating effectively due to 1) a creeping slope above the ditch narrowing or obstructing the ditch, 2) excess vegetation that inhibits efficient water flow, or 3) a profile that forms a sag curve without an apparent culvert to drain the water. The ineffective ditches allow water to migrate under the roadway, which may cause piping which can create voids and a reduction of subgrade strength that can result in pavement damage.

Re-establishing the ditch by cutting the toe of the cut slope can result in increased slope creep.

In order to reduce the ongoing maintenance of cleaning and reestablishing ditches, a wall or buttress, or other mitigation would be required between the cut slope and ditch. However, this mitigation is outside the scope of this project. An underdrain system is recommended to capture groundwater before it migrates under the roadway. Recommendations for underdrain locations are provided in Table B-5.

5. CONSTRUCTION ADVISORY FOR SUBSURFACE STABILIZATION AND

REINFORCEMENT MITIGATION AREAS

5.1 Subsurface Stabilization and Reinforcement Mitigation Limits

The actual limits of many of the subexcavation, deep patch, and MSE wall locations have become obscured as a result of routine maintenance such as patching and overlays. The WFLHD geotechnical engineer can assist the project engineer and contractor in verifying the limits of all subsurface reinforcement locations, if necessary.

5.2 Subsurface Conditions and Temporary Excavations

Most of the subexcavation repair zones have not been investigated, but it is likely that the excavated materials will be similar to what was encountered in the investigation borings.

Encountering scattered cobbles, decomposing organics and voids may be possible. The contractor should anticipate the possible need to fill and compact voids with select borrow prior to constructing subexcavation sections, and removing decomposing organics from below the design subexcavation depth as directed by the CO. Consultation with the project geotechnical engineer may be required.

5.3 Dewatering

The foundation of three of these walls will be at or below lake elevation. The foundation of the wall at Fairholm will be above the lake surface; however, the 3 feet under the foundation will be subexcavated and backfilled with select granular backfill, which will be close to the lake elevation. Some method of dewatering is expected during the construction of the MSE walls. If the excavation is inundated with water, dewatering is required so the backfill can be placed and

Geotechnical Report GR19-15 Page 17 December 2015 compacted in the dry. Extensive dewatering systems including sump and well points may be required to control groundwater infiltration.

The soils within the excavations are anticipated to consist of silty gravel with sand (GM), clayey sand (SC), and silty sand (SM). The maximum allowable temporary slope ratio for excavations less than 20 feet without groundwater infiltration is 1V:1.5 H. The cohesionless soils encountered in the borings will cave-in if left unsupported when excavations extend below the groundwater table. Temporary shoring is anticipated at some of the culvert replacements sites and at the MSE wall locations.

6. LIMITATIONS

This report was prepared to provide final geotechnical recommendations for the Rehabilitate Route 11, Lake Crescent Road (US Highway 101) roadway project. Professional judgments and recommendations are presented in this report. They are based partly on evaluation of the technical information gathered, partly on historical reports and partly on the Geotechnical Team’s general experience with subsurface conditions in the area. It should be noted the borings might not represent potentially unfavorable subsurface conditions between borings.

The observed water levels and/or conditions indicated on the boring logs are as recorded at the time of the exploration. The water levels and/or conditions may vary considerably, with time, according to the climate, rainfall, and other factors and are otherwise dependent on the duration of and methods used in the exploration program.

If, during construction, soil conditions are encountered that vary from those discussed in this report or historical reports, or if design loads, project stationing, and/or configurations change, the Geotechnical Team should be notified immediately in order that it may evaluate effects, if any, on the recommendations contained in this report.

The conclusions and recommendations contained in this report are based on WFL’s Geotechnical Section understandings of the project at the time that the report was written.

7. REFERENCES

Dragovich, J.D., et al (2002) Geologic Map of Washington – Northwest Quadrant, Washington Division of Geologic and Earth Resources, Geologic Map GM-50, Scale 1:250,000.

Federal Highway Administration Lands Highway, 2006, Technical Guidance Manual (TGM), A Geotechnical Supplement to the FLH Project Development and Design Manual.

APPENDIX A

Figures A-1 to A-21: Test Boring Locations on Alignment

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