Halona_Final_Geotech_Report.pdf

Final Geotechnical Report

Halona Bridge Halona Street, Interstate Route H-1 (Adjacent)

Oahu, Hawaii State of Hawaii

Prepared for Federal Highway Administration

Central Federal Lands Highway Division 12300 West Dakota Avenue

Lakewood, CO 80228

January 26, 2016

Hirata & Associates, Inc.

99-1433 Koaha Place

Aiea, HI 96701

9191 South Jamaica Street

Englewood, CO 80112-5946

W.O. 15-5680.1 Page ii

Contents Section Page

1.0 Introduction

1.1 Background and Location

1.2 Purpose and Scope

2.0 Geology and Seismicity

2.1 Regional Geology

2.2 Seismic Design Parameters

3.0 General Subsurface Characterization

4.0 Analysis and Recommendations

4.1 Bridge Foundation

4.1.1 Scour

4.1.2 Predrilling

4.1.3 Pile Axial Resistance

4.1.4 Pile Lateral Load Analysis

4.1.5 Pre-construction Survey and Vibration Monitor

4.1.6 Pile Driving

4.1.7 Test Piles

4.2 Retaining Wall Foundations

4.3 Lateral Design

4.3.1 Active Earth Pressures

4.3.2 Passive Earth Pressures and Coefficient of Friction

4.3.3 Geosynthetic Reinforced Soil Backfill

4.4 Pavement Section

4.5 Site Grading

4.5.1 Site Preparation

4.5.2 Structural Excavations

4.5.3 Onsite Fill Material

4.5.4 Imported Fill Material

4.5.5 Compaction

5.0 Special Contract Requirements

6.0 Additional Services

7.0 Limitations

Appendixes

A Field Exploration Program & Boring Logs

B Laboratory Test Results

C Pile Axial Resistance

D. Boring Logs of Adjacent Bridge of H-1 Freeway

HIRATA & ASSOCIATES, INC.

W.O. 15-5680.1 Page 1

1.0 Introduction

1.1 Background and Location

This report provides geotechnical recommendations for the proposed Halona Bridge replacement located along Halona Street, in the District of Kalihi, Oahu, Hawaii. Halona Bridge is located at MP 20.21, adjacent to the Interstate Route H-1 on-ramp from Vineyard Boulevard. Project vicinity and bridge photos are presented on Figures 1 and 2.

FIGURE 1

Project Vicinity Map

The Central Federal Lands Highway Division (CFLHD) of the Federal Highway Administration (FHWA), in cooperation with the Hawaii Department of Transportation is proposing to replace the existing deficient approximately 103-feet long, 55-feet wide, five span concrete bridge over Kapalama Canal to meet current design standards for roadway width, load capacity, bridge railing and transitions, and bridge approaches.

Based on the Structure Selection Memo, dated March 2015, we understand that the existing bridge abutment walls will remain, but will be cut off at the top. The new bridge abutments will be constructed behind the existing abutment walls, and the replacement bridge will have three spans, and will be about 39 feet wide and about 131 feet long. The abutments and piers of the new bridge will align with the abutments and piers of the adjacent H-1 Freeway Bridge crossing Kapalama Canal. The project also includes new retaining walls adjacent to the H-1 Freeway Bridge, replacing the existing CRM walls.

W.O. 15-5680.1 Page 2

FIGURE 2

Existing Bridge

1.2 Purpose and Scope

This report provides geotechnical recommendations based on engineering analyses of our subsurface investigation, including the general subsurface characterization, seismicity, boring logs, and laboratory test results. Geotechnical recommendations for the design of bridge foundations and site grading are presented in this report.

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2.0 Geology and Seismicity

2.1 Regional Geology

The project site is located in the Kalihi-Palama area on the south side of the Island of Oahu. The Kalihi- Palama area is located in the Coastal Plain geomorphic province of Oahu.

The Island of Oahu is composed largely of weathered remnants of the Waianae and Ko`olau shield volcanoes.

The older Waianae Volcano forms the bulk of the western third of the Island while the younger Ko`olau Volcano forms the majority of the eastern two-thirds of the Island. It is believed that the Waianae Volcano became extinct while the Ko`olau Volcano was still active, and its eastern flank is partially below Ko`olau lavas in Central Oahu.

During the Pliestocene Epoch, sea levels fluctuated in response to the cycles of continental glaciation. As the glaciers grew and advanced, less water was available to fill the ocean basins such that the sea levels fell below the present stands of the sea. When the glaciers melted and receded, an excess of water became available such that the sea levels rose to above the present sea level.

The higher sea level stands caused the formation of deltas and fans of accumulated reef deposits at correspondingly higher elevations, and the deposition of lagoonal/marine sediments in the quiet waters protected by fringing reefs. The lower sea stands caused streams to carve valleys in the sediments and reef deposits.

The project area is located in the alluvial fan area of Kapalama Stream. The surface deposits consist of recent to older alluvium consisting of silts and clays with variable amounts of sand and some pockets of gravel and occasional cobbles. The alluvial soils are underlain by coralline detritus, which originate from fringing coral reef material that was deposited into the stream channel during various stands of sea level.

2.2 Seismic Design Parameters

Based on the AASHTO site classification criteria, the predominant soils at the bridge site located at 21.326o N latitude and -157.867o W longitude can be classified as a stiff soil profile corresponding to a Seismic Site Class D.

Recommended seismic response parameters for use in design based on 7th Edition of the AASHTO LRFD Bridge Design Specifications represents ground motion corresponding to an exceedance probability of approximately 7 percent in 75 years for an earthquake with an approximate 1000-year return period. The 1000-year return period data for the soil at the bridge site was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short and long-period ground acceleration (SS and S1 respectively), and are summarized in Table 1.

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TABLE 1

Summary of Seismic Parameters for Earthquake with 1000-Year Return Period

Horizontal Peak Ground Acceleration, (PGA) 0.174g Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SS) 0.398g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (S1) 0.109g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.45 Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.48 Site Factor at Long-Period Range of Acceleration Spectrum, (FV) 2.36

Factored Horizontal Peak Ground Acceleration, (AS) 0.253g Factored Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDS) 0.590g Factored Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.258g

Seismic Zone Zone 2

W.O. 15-5680.1 Page 5

3.0 General Subsurface Characterization

Table 2 presents a summary of the subsurface characterization. Boring logs and laboratory test results are presented in the Appendixes A and B.

TABLE 2

Summary of Subsurface Materials for Foundation Design Purposes

Subsurface Depth

(ft) Material Description USCS

Classification Apparent Density

Ground Water Depth

(ft)

BORING HA1 (ELEVATION +9.7±) – SW ABUTMENT, HALONA BRIDGE

0.0 to 10.0 Silty SAND with gravel SM Medium dense to loose

8.0

10.0 to 16.5 Silty CLAY CH Soft to medium stiff

16.5 to 87.0 Silty CLAY with sand and gravel CH Stiff to medium stiff

87.0 to 100.5 Silty coralline GRAVEL with sand GM Dense

BORING HA2 (ELEVATION +9.7±) – NE ABUTMENT, HALONA BRIDGE

0.0 to 3.0 Silty GRAVEL GM Medium dense

9.0

3.0 to 18.0 Silty CLAY with gravel CH Medium stiff to soft

18.0 to 28.0 Silty GRAVEL with sand GM Medium dense to dense

28.0 to 100.0 Silty CLAY with sand CH Stiff to medium stiff

100.0 to 109.0 Clayey SILT with coralline sand and gravel ML Stiff

109.0 to 120.5 Clayey SILT with sand MH Medium stiff to stiff

BORING HP3 (ELEVATION -1.0±) – MIDDLE OF CANAL, HALONA BRIDGE

0.0 to 20.0 Silty CLAY with sand CH Soft to medium stiff

(-)1.2

20.0 to 50.0 Clayey SILT with sand and gravel MH Medium stiff

50 to 79.5 Silty CLAY CH Stiff

79.5 to 95 Silty coralline GRAVEL GM Dense

95 to 106.5 Silty CLAY CH Medium stiff

106.5 to 116.0 Clayey SILT with sand MH Stiff

W.O. 15-5680.1 Page 6

4.0 Analysis and Recommendations

This section provides a general interpretation of the site investigation and provides recommendations for design parameters at the proposed structure. Generalized subsurface profiles were developed based on data collected during the investigation. The primary structure considered for this project is replacing the existing five span bridge structure with a three span bridge structure.

4.1 Bridge Foundation

The existing Halona Street bridge was initially supported on piles with assumed pile tip elevation at -12. The bridge was lengthened by one span (about 22 feet) at both ends in 1938. Based on record drawings and discussions with the project Structural Engineer, we understand that the abutments of the lengthened bridge are supported on spread footings, and that strip footings encapsulating the pile tips were constructed for the piers. Based on the record drawings, the bottom of footings are at elevation -12. The abutment wall footings are about 8 feet wide, while the strip footings supporting the piers are about 3.5 feet in width.

The adjacent H-1 Freeway, however, is supported on pile foundations at the Kapalama Canal crossing.

According to record drawings, 16-inch octagonal precast prestressed concrete piles driven to a pile tip elevation of approximately -70 were designed for a vertical pile capacity of 64 kips.

Based on our test borings, the bridge site is underlain by relatively soft and compressible soils at shallow depths, extending to depths of about 16.5 and 18 feet (elevations -7± and -8± respectively) below the roadway surface in borings drilled behind the existing abutments, and about 20 feet (elevation -21±) below canal invert in the boring drilled within the canal.

In order to avoid surcharging the existing abutment walls which will remain and to avoid deep excavations, deep foundations such as driven concrete pile foundations are recommended for support of the new abutments that will be located behind the existing abutment walls.

Based on Section 2 of the 2014 Design Criteria for Bridges and Structures issued by the Hawaii Department of Transportation, the bridge foundation needs to be evaluated against failure using a scour factor of 2.0 (200 percent of the scour) in combination with dead loads of the structure. As a result, based on anticipated scour depths in the range of 8.6 to 9.9 feet depending on pile size (see Table 3 below), a factor of 2.0 will be about 17.2 to 19.8 feet. Therefore, spread footings, if used at the pier locations, will need to be located at depths about 20 feet below the invert of the canal, resulting in extensive shoring and dewatering during construction. Risk associated with dewatering includes potential settlement of adjacent ground and structures caused by drawdown of ground water level.

As a result, deep foundations such as driven concrete piles are recommended to support the proposed replacement bridge at both abutment and pier locations. 16.5 inch octagonal precast prestressed concrete piles were initially recommended. However, we understand that 24-inch octagonal concrete piles are preferred by the project structural engineer. As a results, recommendations for both 16.5-inch and 24-inch concrete piles are presented in this report. In order to reduce the potential distress to the existing canal walls and adjacent residential structures caused by vibration from pile driving, predrilling is recommended for all pile locations. The new piers will be located at the locations of the existing piers and we understand that footings of the existing piers will be removed during construction of the new bridge.

W.O. 15-5680.1 Page 7

Based on the factored ground acceleration coefficient at 1 second, SD1, the bridge site can be classified as Seismic Zone 2 in accordance to Table 3.10.6-1 of the AASHTO LRFD Bridge Design Specifications. Analysis was performed to determine the potential of liquefaction at the bridge abutments. Based on a factored horizontal peak ground acceleration of 0.253 for a 1,000-year return period, and a design earthquake magnitude of 6, factors of safety greater than 1.0 were computed. As a result, it is our opinion that the probability of potential soil liquefaction based on a seismic event with 1,000-year return period is low.

4.1.1 Scour

Scour at the bridge abutments and piers may affect the axial and lateral resistance of the piles. Table 3 presents the total scour at the proposed pier locations provided by the project Hydraulic Engineer on January 15, 2016. It is our understanding that scour will not affect the abutments.

TABLE 3

Design Scour at Pier Locations, 1/15/16

Storm Frequency

Pier

Ground

Elevation (ft.)

16-in Octagonal Piles 24-in Octagonal Piles Scour Depth

(ft.)

Scour Elevation

(ft.)

Scour Depth

(ft.)

Scour Elevation

(ft.)

100-year North -0.70 8.61 -9.31 9.90 -10.60 South -0.10 8.61 -8.71 9.90 -10.00

200-year North -0.70 9.88 -10.58 11.30 -12.00 South -0.10 9.88 -9.98 11.30 -11.40

Based on Section 2 of the 2014 Design Criteria for Bridges and Structures issued by the Hawaii Department of Transportation, 100 percent of the 100-year scour depth will need to be considered in Strength Limit State. 25 percent of the 100-year scour depth will need to be considered in Extreme Event Limit State. In addition, the bridge foundations should be evaluate against failure using 200 percent of the 100-year scour depth will need to be considered in evaluation against failure in combination with dead loads of the structure.

4.1.2 Predrilling

The new abutments and piers will be located relatively close to the existing abutment walls. In order to reduce the potential impact of vibration on the existing walls, predrilling is recommended for all piles locations. The predrilling should extend to a depth of about 15 feet below the bottom of existing wall footings. Based on an elevation of -12 for the existing wall footings, the predrilling should extend to an elevation of approximately -27.

The diameter of the predrilled holes should be slightly smaller than the maximum pile diameter, approximately about 80 to 90 percent of the pile diameter. The annular space between the pile and sidewalls of the predrilled hole, if observed after pile driving, should be backfilled with clean sand, compacted by water jetting or with sand-cement grout.

4.1.3 Pile Axial Resistance

The piles located behind the existing abutment walls should be setback as far behind the existing abutment walls as possible. Based on the proposed bridge length of 131 feet to align with the abutments of the adjacent H-1 Freeway bridge structure, the new abutments are expected to be about 12 to 13 feet behind

W.O. 15-5680.1 Page 8 the existing abutment walls. Based on the subsurface soil conditions encountered, the concrete piles will derive their axial load resistance primarily from friction resistance developed between the perimeter of the pile and the medium stiff to stiff/dense surrounding soils.

Structural loads were not available at the time of this report. Resistance from the surrounding soils above the predrilled depth should be disregarded in determining the pile axial capacity. The subsurface soil below the predrilled elevation of -27 generally consisted of medium stiff to stiff silty clay and clayey silt, with occasional lens of medium dense to dense coralline material. The design of pile axial resistance should follow Section 10.7.3.8.6 of the AASHTO LRFD Bridge Design Specifications. The α-Method and average soil shear strength (Su) of 1.4 ksf may be used to determine the side resistance of the pile. Equation 10.7.3.8.6e- 1 may be used to estimate the tip resistance. The nominal pile axial resistances as a function of pile embedment length below the predrilled depth are presented on Plates C1.1 and C1.2 for 16.5-inch and 24-inch concrete piles respectively. The factored axial resistance can be determined by multiplying the nominal axial resistances by the appropriate resistance factors.

Resistance factors are based on Table 10.5.5.2.3-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and a resistance factor of 0.65 may be used for strength limit state, assuming that either pile dynamic testing (PDA and CAPWAP) or static load test will be performed to confirm the pile axial resistance.

Resistance factor for extreme limit states is 1.0. In general, piles longer than 80 feet in total length will require splicing of the piles. The piles should be spaced a minimum 3 diameters apart, measured from center to center. Structural loads at service limit state were not available at the time of this report.

However, excessive settlement of the friction piles is not anticipated.

4.1.4 Pile Lateral Load Analysis

Lateral resistance of the concrete piles will depend on the stiffness of the surrounding soil, the stiffness of the pile, the boundary condition at the top of the pile, the anticipated unsupported length at top, and the acceptable horizontal displacement of the pile. Computer programs such as LPILE (Ensoft, Inc.) may be used to determine the lateral resistance of the piles. The following soil parameters may be used for the analysis.

TABLE 4

Summary of Material Property for Pile Lateral Load Analysis

Material Description

Unit Weight/ Submerged Unit Weight

(pcf)

Cohesion (psf)

Strain Factor ( ε50)

Friction Angle (deg)

Subgrade Modulus

(k) Loose to Medium Dense Silty Sand 115/55 - - 30 Default

Soft to Medium Silty Clay 110/50 600 Default - -

Medium Stiff to Stiff Silty Clay/Clayey Silt 115/55 1400 Default - -

We recommend that the default values provided by LPILE be used for the strain factor (ε50) and initial subgrade modulus (k) in the analysis. For piles supporting the piers, resistance from soil layers above the scour depth should be disregarded when scour effects need to be considered.

The potential effects of pile group interaction should follow recommendations based on Table 10.7.2.4-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications. Based on Table 10.5.5.2.3-1 in the 7th

W.O. 15-5680.1 Page 9

Edition of the AASHTO LRFD Bridge Design Specifications, the applied strength limit state resistance factor is 1.0.

Since the piles supporting the new abutments are expected to be more than 5 to 6 pile diameters behind the existing abutments walls, lateral movement of the piles in the longitudinal direction toward the walls is expected to impose negligible pressure on the walls. We recommend that a minimum horizontal distance of 5 pile diameters be maintained between the concrete piles and the back face of the existing abutment wall.

4.1.5 Pre-construction Survey and Vibration Monitor

Prior to pile driving, we recommend that the contractor perform a survey of all major existing structures within 75 feet of the abutment and pier locations. All visible defects, cracks, anomalies and issues that could be affected by the pile driving activities should be documented and photographed.

Vibration due to pile driving could affect the adjacent structures. Structures within 100 feet of the pile driving should be monitored for vibration during pile driving by the contractor.

4.1.6 Pile Driving

Selection of pile hammer will depend on the pile length and design pile capacity. Although the final pile length and design pile capacity were not available at the time of this report, based on past experience, it is our opinion that the piles should be driven with a hammer capable of delivering energies in the range of 20,000 to 28,000 foot-pounds of energy per blow for the 16.5-inch octagonal concrete piles. Larger hammer delivering higher energies, in the range of 26,000 to 41,000 foot-pounds per blow, may be required for the 24-inch diameter piles. To prevent development of excessive stresses in the piles during the driving operations, the hammer energy (or stroke) should be carefully controlled during the initial stages of pile driving. The minimum driving resistance required will depend on the type of hammer and pile driving equipment used. The contractor should conduct drivability studies to select hammers that will not cause the driving stresses (compression and tension) in the pile to exceed acceptable levels.

4.1.7 Test Piles

Indicator test piles should be driven to help determine production pile lengths and pile driving criteria. In general, approximately 10 percent of the total number of piles for the project should be driven during the indicator test pile driving operations. A minimum of 2 test piles are recommended at each abutment and pier locations.

Indicator test piles should be at least 10 feet longer than that estimated for production piles and should be driven with the same hammer which will be used for production pile driving. The test driving should be monitored by an engineer from our staff so that any necessary changes due to unexpected conditions encountered during the test driving operations can be evaluated.

The indicator test piles should be monitored with a pile driving analyzer (PDA) to confirm the design axial capacity of the piles and the pile driving criteria. The dynamic testing of the piles should be performed in general accordance with ASTM D 4945. A Signal Matching program analysis should be performed to confirm the load capacity of the piles at the end of the test driving program.

Since PDA and Signal Matching program analyses are not readily available in Hawaii, as an alternative to PDA of the indicator test piles and Signal Matching program analyses, a static load test could also be perform to confirm the design axial capacity of the piles. The static load test should be performed on one of the indicator test piles. The load test pile should be subjected to at least 100 percent of the nominal design pile

W.O. 15-5680.1 Page 10 bearing capacity. The location of the load test pile should be determined by the construction officer after the indicator test pile program during construction.

4.2 Retaining Wall Foundations

Record drawings indicate that bottom of existing abutment wall footings are at elevation -12, with a footing width of 8 feet. Based on our borings, the footings are expected to be founded on medium stiff to stiff silty clay and/or medium dense to dense silty gravel. As a result, the following bearing values may be used for evaluation of the existing abutment wall foundations. The bearing resistance factors are based on Table 10.5.5.2.2-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and the applied strength limit state resistance factor for bearing resistance is 0.45.

TABLE5

Bearing Value for Existing Abutment Wall Foundations

Extreme Event Limit State 7,800 psf

Strength Limit State 3,500 psf

Service Limit State 2,600 psf

The footing elevations for the new retaining walls adjacent to the H-1 Freeway Bridge were not available at the time of this report. Based on discussion with the project structural engineer, the bottom of footings are expected to be at about elevation +3. Based on our borings, the footings are expected to be founded on either the soft to medium stiff silty clay, or the backfill material after removal of the existing CRM walls. Due to the anticipated soft to medium stiff condition of the subgrade soil, additional site preparation work is recommended for the footings. We recommend that bottom of footing excavations be overexcavated to a minimum depth of about 3 feet and replaced with compacted granular structural fill enveloped in a geotextile filter fabric. If the overexcavations extend below groundwater, the backfill below groundwater level should consist of clean gravel. The clean gravel should be thoroughly tamped with vibratory equipment or other construction equipment to an unyielding condition. The clean gravel should be separated from the overlying granular structural fill by the geotextile filter fabric. The granular structural fill should be compacted on a minimum 95 percent compaction as determined by AASHTO T-180 (ASTM D 1557).

The following bearing values may be used to design the retaining wall foundations founded on a minimum 3 feet of compacted granular structural fill/clean gravel. The relatively low bearing values are recommended to limit the potential settlement of the retaining wall foundations.

TABLE 6

Bearing Value for New Retaining Walls Adjacent to H-1 Freeway Bridge

Extreme Event Limit State 4,500 psf

Strength Limit State 2,100 psf

Service Limit State 1,500 psf

4.3 Lateral Design

The following earth pressures may be used to evaluate the existing abutment walls and for design of new retaining walls.

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Based on our test borings, soils behind the existing abutment walls consisted of medium dense to loose silty sand and medium stiff to soft silty clay. The following soil parameters were used to determine the earth pressures for design. A friction angle of 28 degrees is determined for the silty clay based on correlations with Plasticity Index of the soils. The depth to groundwater can be expected to vary with tidal fluctuations and with water level within the canal. A high groundwater level of elevation +2 may be assumed for design.

Global slope instability of the existing abutment walls is not anticipated.

TABLE 7

Summary of Material Property for Lateral Earth Pressure Design

Material Description

Unit Weight/ Submerged Unit Weight

(pcf)

Cohesion

(psf)

Friction Angle

(deg) Loose to Medium Dense

Silty Sand 115/55 0 30

Soft to Medium Stiff Silty Clay 110/50 0 28

4.3.1 Active Earth Pressures

Active earth pressures were computed using the Coulomb Criteria. Based on the computed values and local practice, equivalent fluid soil unit weights of 40 and 16 pounds per cubic foot are recommended for above and below groundwater respectively. In general, rotation at the top of the wall or lateral translation equals to or greater than 0.002 times the wall height may be required to reach the active pressure condition.

For dynamic lateral earth pressure considerations, the Mononobe-Okabe (M-O) pseudo-static theory can be utilized to estimate the dynamic lateral earth pressure on retaining walls due to earthquake effects. For walls that are free to move laterally up to 1 to 2 inches or rotate in the event of an earthquake, a horizontal acceleration coefficient equivalent to ½ times the maximum peak ground acceleration may be assumed. The vertical acceleration coefficient can be neglected when calculating lateral loads. A resistance factor of 1.00 should be assumed for the extreme event limit state.

4.3.2 Passive Earth Pressures and Coefficient of Friction

Passive earth pressures were computed using the Log-Spiral Method. A wall friction angle value equal of 1/2 times the soil’s friction angle was used. Based on the computed values and local practice, the following passive earth pressures are recommended. Resistance factors are based on Table 10.5.5.2.2-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications. The applied strength limit state resistance factor for passive pressure resistance is 0.5. Unless covered by pavement or concrete slabs, the upper 12 inches of soil should not be considered in computing lateral resistance.

TABLE 8

Passive Pressures

Soil Type Extreme Event Limit State (pcf)

Strength Limit State (pcf)

Above groundwater 470 235

Below groundwater 210 105

W.O. 15-5680.1 Page 12

The lateral loads imposed on the shallow wall foundation due to lateral earth pressure and horizontal seismic forces may also be resisted by the friction developed between the wall footing and the supporting soils with a resistance factor of 0.85. A nominal sliding resistance of 0.53 may be assumed.

4.3.3 Geosynthetic Reinforced Soil Backfill

In order to reduce the earth pressures on the existing abutment retaining walls, we understand that Geosynthetic reinforced soil (GRS) backfill is being considered to replace the upper 4 to 6 feet of existing abutment retaining wall backfill.

Geosynthetic reinforced soil (GRS) backfill consists of closely spaced layers of geosynthetic reinforcement, typically 12 inches or less, and compacted granular fill material. The geosynthetic reinforcement are similar to tieback elements that provide tensile resistance to driving forces, while the closely spaced nature of geosynthetic reinforcement creates a confined soil mass such that the GRS mass is essentially freestanding and self-supporting. GRS backfill is expected to exert very little or no earth pressure on the walls in front of it.

Design of the GRS backfill was based on methods presented in NCHRP Document 187 – Seismic Design of Geosynthetic Reinforced Soil Bridge Abutments with Modular Block Facing, and FHWA-HRT-11-026 – Geosynthetic Reinforced Soil Integrated Bridge System. Based on our analyses for a GRS backfill height of about 4 to 6 feet , we recommend that the geosynthetic reinforcements be spaced at a vertical distance of 12 inches or less and extend horizontally to a minimum distance of 6 feet behind the GRS front face. An overlapping length of 3 feet may be used to wrap the reinforced soil layers around the front face the backfill and embedded into the backfill.

Based on FHWA-HRT-11-026, a biaxial, woven polypropylene geotextile with a minimum ultimate strength of 4,800 pounds per foot is recommended. In addition, the geotextile should have a minimum long term allowable design strength of 1,000 pounds per foot. The geotextile should conform to FP-14 714.04, Type IV or greater.

The reinforced fill material should consist of aggregate base course material. The aggregate base course should conform to FP-14 703.05, Gradation Designation C (Base). The aggregate base course should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 95 percent compaction as determined by AASHTO T-180 (ASTM D 1557).

As a precautionary measure, an expanded polystyrene geofoam (EPS), about 6 to 12 inches thick, may also be placed between the GRS mass and the back face of the existing abutment retaining walls to further reduce the potential effects of the lateral soil displacements of the GRS backfill from surcharging the walls during a seismic events. EPS Geofoam, ASTM C578 Type VIII or equivalent is recommended.

4.4 Pavement Section

The AC concrete pavement encountered in borings HA1 and HA2 consisted of approximately 7 and 4 inches of asphaltic concrete, respectively, over approximately 12 inches of base material.

Due to the relatively short length of pavement for bridge approaches, based on discussions with Central Federal Lands Highway Division, we understand that new pavement will generally match the existing pavement sections identified in the borings.

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4.5 Site Grading

4.5.1 Site Preparation

The project site should be cleared of all concrete and AC pavements, and other deleterious material. Prior to fill placement, the existing ground should first be scarified to a depth of six inches, moisture conditioned to about 2 percent above optimum moisture content, and compacted to a minimum 90 percent compaction as determined by AASHTO T-180 (ASTM D 1557). Soft or loose soils, indicated by pumping conditions, should be removed and replaced with imported granular structural fill.

4.5.2 Structural Excavations

Based on our exploratory test borings, we believe that excavations into the surface silty sand, silty gravel, and silty clay below the roadway surface can be accomplished with conventional excavating equipment.

Temporary cuts should be stable at slope gradients of 1H:1V (horizontal to vertical) or flatter for temporary conditions. Shoring may be required for excavations extending into the soft/loose deposits below groundwater.

Shoring and dewatering will be required for excavations to remove existing pier footings within the canal. It should be the Contractor’s responsibility to conform to all OSHA safety standards for excavations.

Dewatering for construction is the responsibility of the contractor, and the selection and method of dewatering should be determined by the contractor. The dewatering method selected should be designed to have minimal impact on the groundwater level surrounding the project site. The contractor should address the potential for settlement of adjacent structures and the monitoring effort that will be implemented to detect ground movement or settlement of existing structures and infrastructure in his dewatering program.

4.5.3 Onsite Fill Material

The silty sand and gravel above groundwater table may be reused in compacted fills and backfills, provided all rock fragments larger than 3 inches in maximum dimension are removed. The onsite silty clay should not be reused in compacted fills and backfills.

4.5.4 Imported Fill Material

Imported structural fill should be well-graded, non-expansive granular material and conform to FP-14 704.04 with modifications presented in 5.0 Special Contract Requirements section of this report.

4.5.5 Compaction

Structural fill, granular backfill and granular embankment fill should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 95 percent compaction as determined by AASHTO T-180 (ASTM D 1557). Fill consisting of cohesive soils such as the onsite silty clay should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 90 percent compaction.

Fill placed in areas which slope steeper than 5H:1V should be continually benched as the fill is brought up in lifts. Fill placed on slopes should be keyed and benched into the existing slope to provide stability for the new fill against sliding.

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5.0 Special Contract Requirements

Section 261. – Reinforced Soil Slopes and Geosynthetic Reinforced Soil (GRS) Backfill

Description

261.01 This work consists of constructing reinforced soil slopes and GRS backfill.

Material

261.02 Conform to the following Subsections:

Backfill material 704.03 GRS Backfill material 703.05 Reinforcement geotextile and geogrid 714.04

Expanded Polystyrene (EPS) Geofoam 725.19

Construction Requirements

261.03 Submittals. At least 30 days before starting reinforced soil slope and GRS backfill construction, submit the following according to Subsection 104.03:

(a) Manufacturer’s certification. Certification and test results indicating the proposed soil reinforcement, reinforced embankment, and facing material meets requirements;

(b) Splice details. Detail proposed splices in soil reinforcement; and

(c) Facing details. Provide installation methods and connection details for facing system.

261.04 General. See Section 257 for Contractor-designed reinforced soil slope systems.

Verify the limits of the reinforced soil slope and GRS backfill installation.

Clear the area of vegetation and obstructions according to Sections 201 and 203.

Conserve topsoil according to Subsection 204.05.

Prepare and compact foundation soils according to Subsection 204.09(a). Grade the foundation soils supporting the reinforced soil slope and GRS backfill for a width equal to the length of the lowest soil reinforcement level. Bench cut slope surfaces to key the reinforced soil slope and GRS backfill into the existing embankment.

261.05 Soil Reinforcement and Facing. Place the bottom layer of reinforcement so that it is in full contact with the level subgrade and so that it extends from the rear of the abutment or slope face to the limits shown on the plan drawings. Ensure that no voids exist beneath the reinforcement. Place the reinforcement so that the strongest direction is perpendicular to the abutment of slope face.

Lay each layer of soil reinforcement flat, Pull the reinforcement tight, and hold in place with pins, soil piles, or other approved methods. Construct the reinforced soil slope and GRS backfill to achieve the slopes specified and roadbed tolerances according to Subsection 204.13(d).

W.O. 15-5680.1 Page 15

Wrap the reinforcement as shown on the plan drawings. When seams are necessary, stagger at least 24-inches apart on subsequent lifts. Seams are only allowed in the direction perpendicular to the abutment or slope face. Seams of adjacent sheets of reinforcement shall overlap a minimum of 12 inches. Except at the wrapped face of the slope or abutment, an approximately 1-inch thick layer of backfill shall be placed between the two overlapping reinforcement sheets.

Grade and compact backfill material according to Subsections 204.09 and 204.11 in maximum 8-inch loose lifts before placing the next soil reinforcement layer. No equipment is allowed directly on the reinforcement. Maintain a minimum cover of 6 inches (150 millimeters) over the reinforcement during spreading and compacting of fill material. Avoid sudden stops, starts, or turns of the equipment. Do not use sheepsfoot-rollers. Use lightweight mechanical tampers, rollers, vibratory systems, or other methods for compaction within 36 inches (900 millimeters) of the slope face.

Place backfill material in a manner that tensions the reinforcement working from the slope face or GRS backfill face to the back of reinforcement. Remove any wrinkles that form during backfilling operations.

Do not damage or disturb the reinforcement elements. Do not wrinkle, tear, or displace the reinforcement. Carefully correct all damaged, misaligned, or distorted reinforcement elements.

Fill and compact any depressions caused during compaction so that the compacted lift is level prior to constructing the next layer.

Do not leave reinforcement exposed at end of shift. At the end of each shift, shape to drain and compact the work area to a uniform cross-section. At the end of the day’s operation, slope the last lift of backfill away from the slope face to direct surface runoff away from the slope.

261.06 Acceptance. Material for reinforced soil slopes and GRS backfill will be evaluated under Subsections 106.02, 106.03, and 106.04.

Construction of reinforced soil slopes will be evaluated under Subsections 106.02 and 106.04.

Excavation will be evaluated under Section 204.

Measurement

261.07 Measure the Section 261 pay items listed in the bid schedule according to Subsection 109.02.

When measuring reinforced soil slope and GRS backfill within the stabilized volume by the cubic yard (cubic meter), measure in place.

Payment

261.08 The accepted quantities will be paid at the contract price per unit of measurement for the Section 261 pay items listed in the bid schedule. Payment will be full compensation for the work prescribed in this Section. See Subsection 109.05.

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Section 551. – Driven Piles

551.03 Qualifications.

(a) Professional engineer. Delete the text and substitute the following:

Name of Hawaii licensed civil engineer specializing in geotechnical engineering and a resume describing at least 3 years’ experience in wave equation analyses to perform the wave equation analysis.

551.06 Pile Lengths. Delete the Subsection and substitute the following:

551.06. Pile Lengths. Furnish piles with sufficient length to obtain the required resistance and to extend into the pile cap or footing as indicated in the plans. Furnish test piles at least 10 feet longer than the estimated length of production piles. In addition, increase the length to provide fresh heading and to provide for the Contractor’s method of operation. When test piles are required, furnish piles in the lengths determined by the test piles.

551.07 Test Piles. Delete the third paragraph and substitute the following:

Excavate the ground at the site of each test pile or production pile to the elevation of the bottom of the footing before the pile is driven, unless otherwise accepted by the CO. Furnish test piles at least 10 feet longer than the estimated length of production piles. Drive test piles with the same equipment as the production piles.

551.09 Preboring. Delete the Subsection and substitute the following:

551.09 Preboring. Use auguring, wet rotary drilling, or other approved methods of predrilling.

Prebore all pile locations to the depths specified in the plans, or to elevation -27, or as directed by CO.

Prebore holes slightly smaller than the diameter or diagonal of the pile cross-section while allowing penetration of the pile to the specified predrill depth. Increase the hole diameter to the least dimension adequate for pile installation if the subsurface obstruction such as boulders or rock layers are encountered.

Fill remaining void space around the pile with sand, grout, or other approved material after driving is complete. Do not use a punch or a spud instead of preboring.

Do not impair the capacity of existing piles or the safety or condition of adjacent structures. If preboring disturbs the capacity of previously installed piles or structures, restore the required nominal capacity of piles and structures by approved methods.

551.10 Preparation and Driving. Add the following after the first paragraph of the subsection:

Perform preconstruction condition survey of critical structures, such as existing canal walls and residential structures, within 75 feet radius of pile locations prior to pile driving, take photographs, video tapes, and/or notes of any existing structural, cosmetic, plumbing or electrical damage/cracks. Install crack displacement monitoring gages as appropriate across any significant existing cracks to help verify any additional distress if it should develop. Take readings of the gages prior to pile driving, as well as during the pile driving activities.

Employ a qualified vibration specialist to establish a safe vibration level and monitor the vibrations during pile driving activities. Adjust pile driving energy if necessary to stay within safe vibration level.

551.13 Pile Load Tests. Delete the first paragraph and substitute the following:

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Either perform dynamic load tests on all test piles or conduct static load test on a test pile selected by CO.

(b) Static load tests.

(1) Delete the text and substitute the following:

Have a Hawaii licensed structural engineer prepare the drawings;

Section 704.04 Structural Backfill. Furnish a suitable, well-graded, non-expansive granular material conforming to the following:

(a) Maximum particle size 3 in (75 mm)

(b) Material passing No. 200 (75-µm) between 8 and 20 percent sieve, AASHTO T 27 and AASHTO T 11

(c) Plasticity index, AASHTO T 90 10 max.

(d) California Bearing Ratio, AASHTO T 193 CBR value, 15 min.

(ASTM D 1883) CBR expansion value, 1.0 percent max.

Section 725.19 Expanded Polystyrene (EPS) Geofoam (Added Subsection)

Furnish polystyrene foam board, ASTM C578 Type VIII or equivalent.

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6.0 Additional Services

We recommend that we perform a general review of the final design plans and specifications. This will allow us to verify that the foundation design and earthwork recommendations have been properly interpreted and implemented in the design plans and construction specifications.

For continuity, we recommend that we be retained during construction to (1) observe all pile driving work, including predrilling, indicator test pile driving, and static load testing, (2) observe footing excavations prior to placement of reinforcing steel and concrete, (3) review and/or perform laboratory testing on import borrow to determine its acceptability for use in compacted fills, (4) observe structural fill placement and perform compaction testing, and (5) provide geotechnical consultation as required.

Our services during construction will allow us to verify that our recommendations are properly interpreted and included in construction, and if necessary, to make modifications to those recommendations, thereby reducing construction delays in the event subsurface conditions differ from those anticipated.

Appendix A Field Exploration Program & Boring Logs

Plate A1.1

Appendix A: Field Exploration Program & Boring Logs

A1.0 Introduction Hirata & Associates, Inc. completed a field exploration program for Halona Bridge, Halona Street, Interstate Route H-1 (Adjacent), District of Kalihi, Island of Oahu, State of Hawaii, Task Number 499074.02.SU.HI, between September 15, 2014, and January 23, 2015 for the new bridge. The scope of work for the field exploration program included drilling two borings along Halona Street at the rear of existing bridge abutments and one boring in the Kapalama Canal adjacent to the bridge. Each component of the field exploration program was coordinated and observed by a Geotechnical Engineer from Hirata & Associates, Inc. Boring location plans, as well as individual boring logs are attached. These logs represent a compilation of field data and description of the soil samples by Hirata & Associates, Inc. personnel. The methods used to conduct the field exploration program are described below. Representative soil and rock samples collected during the field exploration program were transported to Hirata & Associates, Inc. in Aiea, Oahu, Hawaii for testing. A summary of the laboratory testing program is provided in Appendix B.

A2.0 Explorations Hirata & Associates, Inc. of Aiea, Oahu, Hawaii provided the drilling services for the soil borings. Borings were completed using a Mobile B80 truck-mounted drill rig and portable drilling equipment.

The initial phase of borings was drilled between September 15 through 19, 2014. Two borings at the rear of existing bridge abutments, HA1 and HA2, were completed to depths of about 100.5 and 120.5 feet below ground surface to explore subgrade conditions. One boring within the Kapalama Canal adjacent to the existing bridge was drilled between January 13 and 23, 2015. The boring, HA3, was completed to a depth of about 116 feet below the invert of the canal. Borings were drilled and sampled using continuous flight solid stem augers and continuous sample rock coring techniques.

The locations of individual borings were located in the field by measuring/taping offsets from existing features shown on the plans. Boring locations are shown on the Boring Location Plan, Plate A2.2, and are shown below in Table A-1. Surface elevations at boring locations were estimated based on the Topographic Survey provided by CH2M Hill received on December 2, 2014. Following drilling activities, the borings were backfilled with cuttings generated during the drilling operations and, where applicable, capped with “Instant Road Repair” by International Roadway Research.

TABLE A-1

Field Exploration Locations

Foundation Designation Boring

Number Station Offset Ground

Elevation Termination

Elevation SW Abutment HA1 10+96 8 LT 9.7 ft -90.8 ft NE Abutment HA2 12+48 7 RT 9.7 ft -110.8 ft

Piers HA31 12+00 35 RT -1.0 ft -117 ft

A3.0 Soil and Rock Sampling Representative samples were obtained using Shelby tubes in general accordance with ASTM D 1587 and a modified California (MC) barrel sampler in general accordance with ASTM D 3550. The MC test procedure involves driving a 3-inch outside diameter sampler lined with 2.42-inch-diameter brass sleeves a depth of 18

Plate A1.2 inches with a 140-pound hammer falling a distance of 30 inches. The number of blows required to advance the sampler through each of the 6-inch increments was recorded. The MC resistance, or 3-inch N-value, is defined as the number of blows required to drive the sampler over the second and third 6-inch increments.

The N-value provides a means for evaluating the relatively density or compactness of cohesionless (granular) soils and consistency or stiffness of cohesive (fine-grained) soils. Because high penetration resistance prevented driving the total length of the sampler at times, the penetration resistance for the partial penetration was recorded where applicable. As a result of the larger diameter barrel sampler compared to the Standard Penetration Test (SPT) sampler, the MC sampler yields slightly higher raw blow counts when compared to SPT N-values for similar soils. The raw MC blow counts over the second and third 6-inch increments were used to define the relative density and consistency of the subsurface materials. The Shelby and MC samples were sealed with plastic end caps and transported to Hirata & Associates, Inc. for testing.

An attempt was made to collect a core sample of the silty coralline gravel from boring HA3 using a NX core barrel having an inside diameter of 2.1 inches. No recovery was obtained from the core run.

A4.0 Soil Classification System During the drilling of borings, Hirata & Associates, Inc. personnel collected soil samples and prepared field logs of the borings. Soil identification and descriptions, as shown on the field logs, are based on ASTM D2488, a systematic process for identifying and describing individual soil samples by visual and manual means. When sufficient laboratory testing was completed, select samples from borings were classified using the Unified Soil Classification System (USCS). Both the visual soil identification system and the referenced soil classification systems are summarized in the attached Boring Log Legend and USCS plates.

Appendix B Laboratory Test Results

Plate B1.1

Appendix B: Laboratory Test Results

B1.0 Introduction Laboratory tests were completed on select soil samples recovered from the field exploration program in general accordance with ASTM testing methods. The laboratory testing program was completed to provide data for engineering studies and to classify the materials into similar geologic groups. The testing program included index tests and geotechnical engineering property tests. The following sections describe the laboratory testing procedures and test results are presented in the attached laboratory reports.

B2.0 Index Tests Classification and index laboratory testing included identification by visual and manual means, and tests to determine natural water content, unit weight, grain size distribution, and Atterberg limits. When sufficient laboratory testing was completed, select samples from borings were classified using the Unified Soil Classification System (USCS).