2-9 -Geotechnical Eng Report.pdf

SUBMITTED TO:

Brown and Caldwell, Inc.

6500 S Macadam Ave Suite 200 Portland, OR 97239

BY:

Shannon & Wilson 3990 Collins Way, Suite 100 Lake Oswego, OR 97035

(503) 210-4750 www.shannonwilson.com

GEOTECHNICAL ENGINEERING REPORT

Vancouver Water Station 14 PFAS Treatment Design

VANCOUVER, WA

July 2024

Shannon & Wilson No: 112510

Vancouver Water Station 14 PFAS Treatment Design Geotechnical Engineering Report

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CONTENTS

1 Introduction

1.1 Project Understanding and Site Description

1.2 Site Description

1.3 Scope of Services

2 Geologic Setting

2.1 Seismic Setting

2.1.1 Cascadia Subduction Zone: Mega-Thrust Interface Source

2.1.2 Cascadia Subduction Zone: Intraslab Source

2.1.3 Shallow Crustal Source

3 Subsurface Exploration Program

4 Laboratory Testing

4.1 Corrosivity Testing Results

5 Subsurface Conditions

5.1 Geotechnical Units

5.2 Groundwater

5.3 Seismic Design Ground Motions

5.4 Liquefaction

5.4.1 General

5.4.2 Liquefaction Analysis and Liquefaction-Induced Settlement

5.5 Lateral Spreading

5.6 Fault Surface Rupture

6 Geotechnical Engineering Recommendations

6.1 General

6.2 Foundation Design – Tanks and Vessels

6.2.1 Tank and Vessel Settlement

6.2.2 Tank and Vessel Lateral Resistance

6.3 Foundation Design – Restrooms/Storage Building

6.3.1 Allowable Bearing Capacity – Restroom/Storage Building

6.3.2 Building Foundation Settlement

6.3.3 Foundation Lateral Resistance – Restroom/Storage Building iii

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6.4 Lateral Earth Pressures on Buried Structures and Retaining Walls

6.4.1 Static and Seismic Lateral Earth Pressures

6.4.2 Embedded Wall Drainage

6.4.3 Surcharge Loading

6.4.4 Lateral Resistance

6.5 Pavement Design

7 Construction Considerations

7.1 Earthwork

7.2 Subgrade Preparation

7.3 Wet Weather Construction

7.4 Reinforced Crushed Rock Mat

7.5 Structural Fill Placement and Compaction

7.6 Excavation and Groundwater Control

7.7 Pavement Materials

8 Limitations

9 References

Exhibits Exhibit 2-1: USGS Class A Faults Within an Approximate 15-Mile Radius of the Project Site .6 Exhibit 5-1: Depth to Groundwater Summary Exhibit 5-2: USGS Code-Based MCE and Design Seismic Parameters Exhibit 5-3: Summary of Liquefaction-Induced Settlement at Depth Exhibit 6-1: Passive Equivalent Fluid Pressures Exhibit 6-2: Recommended Pavement Thicknesses

Figures Figure 1: Vicinity Map Figure 2: Site and Exploration Plan Figure 3: Lateral Earth Pressure Distribution on Permanent Embedded Walls Figure 4: Recommended Surcharge Loading for Temporary and Permanent Walls Figure 5: Typical Reinforced Crushed Rock Mat

Appendices Appendix A: Field Explorations iv

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Appendix B: Laboratory Testing Important Information v

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ACRONYMS

ASCE American Society of Civil Engineering ASTM American Society for Testing and Materials bgs Below Existing Ground Surface CSZ Cascadia Subduction Zone GAC Granular Activated Carbon IBC International Building Code MCE Maximum Considered Earthquake pci pounds per cubic inch pcf pounds per cubic foot PFAS Per- and Polyfluoroalkyl substances PGA Peak Ground Acceleration psf Pounds Per Square Foot SPT Standard Penetration Test USGS United States Geological Survey VFD Variable Frequency Drive VWP Vibrating Wire Piezometer DCP Dynamic Pone Penetrometer Mm/yr millimeters per year Mw moment magnitude CSZ Cascadia Subduction Zone SPT Standard Penetration Test CMU Cement Modular Unit Mr Resilient Modulus

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1 INTRODUCTION

1.1 Project Understanding and Site Description

We understand the Vancouver Water Station 14 (WS14) PFAS Treatment Design project (the Project) is proposed to address water quality at WS14. In late 2020, per- and polyfluoroalkyl substances (PFAS) were first detected in the groundwater at the station. We understand the Project includes the preliminary design of an onsite PFAS treatment facility at WS14.

The proposed PFAS treatment facility will include three pairs (6 total) of granular activated carbon (GAC) treatment vessels supported on an at-grade structural slab foundation. The at-grade slab will have an approximate footprint of 114 feet by 32 feet and will have a top of slab elevation of 270.5 feet. The tallest point on the proposed GAC vessels will be on the order of 30 feet above ground and accordingly, the proposed GAC vessels are anticipated to have a fundamental period of vibration less than 0.5 seconds.

The proposed Model 12-40 GAC's are 12-feet in diameter and 26 feet and 9 inches in height.

Each GAC vessel is specified to have a system operating weight of 385,000 pounds.

Backwash waste from the GAC will be sent to an 80,000-gallon backwash tank north of the PFAS treatment facility. An approximate 70-foot by 25-foot infiltration area for backwash waste was originally considered immediately north of the backwash tank; however, this has been removed from the proposed upgrades.

Other site improvements include below grade, precast concrete vaults, and a new paved driveway to allow service trucks to access the new PFAS treatment facility and be able to complete a full turnaround. In addition, a new at-grade restroom/storage facility is proposed north of the backwash tank. The restroom/storage facility will have a finish floor elevation of 270.0 feet and we assume will have wall loads less than 1 kip per lineal foot.

The site grading plan calls for a small interlocking block retaining wall (less than 4 feet in height) on the SW corner of the site as shown on Figure 2.

1.2 Site Description

WS14 is located southwest of the intersection of NE 78th Street and NE Andresen Road in Vancouver, Washington. The site currently consists of 3 groundwater wells (Wells #1 through #3) that feed into an aeration tower and subsequent wet well. A single-story treatment building is located in the southwest portion of the site which treats water pumped from the wet well via two VFD driven booster pumps. A paved driveway exists along the western margin of the site providing access to the treatment building and well houses.

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A grove of large trees is located in the proposed PFAS treatment facility located east of Well #2 and the existing treatment building. An existing wetland area is north of the proposed PFAS treatment facility.

1.3 Scope of Services

Shannon & Wilson's services were conducted in general accordance with the scope of services specified in our Task Order Authorization with Brown and Caldwell dated October 1, 2023. The scope of services performed for this study consisted of the following tasks:

Review existing information, including available geologic maps and previous geotechnical reports;

Complete two borings to depths of 71.5 feet below ground surface (bgs);

Complete two test pits to depths between 3.0 and 6.6 feet bgs;

Complete two dynamic cone penetrometer (DCP) tests to support pavement design;

Complete one infiltration test at a depth of 3.0 feet bgs;

Conduct laboratory testing following the completion of the subsurface explorations;

Provide code based seismic ground motion parameters in accordance with 2022 Oregon Structural Code and ASCE 7-16;

Provide geotechnical recommendations for shallow foundation support of the new PFAS treatment facility, including estimates of allowable bearing capacity, total static and differential settlement;

Provide recommendations for pavement design;

Evaluate geologic hazards at the site including liquefaction, lateral spreading, and surface fault rupture;

Provide construction considerations including subgrade preparation, reuse of on-site materials, and preparation and compaction of site fills and backfill; and

Prepare this geotechnical engineering report.

2 GEOLOGIC SETTING

The geology of the project area has been extensively mapped (Wells and others, 2018, 2020, O’Connor and others, 2016) as Pleistocene (between 2.6 million to 11,000 years ago) catastrophic flood deposits. Holocene and Pleistocene (2.6 million years ago to present) Eolian deposits are also mapped within the project area.

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The Eolian deposits are described as massive sand and silt deposits broadly covering upland areas. Most upland areas in the project vicinity are covered by variable thicknesses of eolian deposits, however, are difficult to distinguish from catastrophic flood deposits (O’Connor and others, 2016).

Pleistocene (between 2.6 million to 11,000 years ago) glacial-outburst flood deposits underlay Eolian deposits. The glacial flood deposits typically consist of sand and gravel sediments. The flood sediments were transported and deposited by as many as 40 or more very large, catastrophic floods that originated in western Montana. During the late stages of the last great ice age, between about 18,000 and 15,000 years ago, a lobe of the continental ice sheet repeatedly blocked and dammed the Clark Fork River in western Montana, which then formed an immense glacial lake called Lake Missoula. The lake grew until its depth was sufficient to buoyantly lift and rupture the ice dam, which allowed the entire massive lake to empty catastrophically. Once the lake had emptied, the ice sheet again gradually dammed the Clark Fork Valley and the lake refilled, leading to 40 or more repetitive outburst floods at intervals of decades (Allen and others, 2009). During each short-lived episode, floodwaters washed across the Idaho panhandle, through the eastern Washington scablands, and through the Columbia River Gorge. Because of constrictions in the drainage path, the floodwaters backed up and formed lakes into which sediments were deposited.

One such lake covered parts of Clark County including the project site, and extended south to Eugene, Oregon. The flood deposits can be as thick as 200 feet (O’Connor and others, 2016).

2.1 Seismic Setting

The contemporary tectonics and seismicity of the region are the result of oblique, northeastward subduction at a rate of about 37 millimeters per year (mm/yr) (DeMets and others, 2010) of the Juan de Fuca oceanic plate beneath the North American continental plate (Wells and others, 1998; Wells and Simpson, 2001). This complex tectonic setting produces east-west compressive strain along the Cascadia Subduction Zone (CSZ), as well as northward translation and rotation of the mobile, crustal, Cascadia forearc blocks that span the leading edge of the North American plate (Wells and others, 1998; McCaffrey and others, 2007, 2013). Rotation of the Sierra-Nevada block and expansion of the Basin and Range drive the northward migration and clockwise rotation of the Cascadia forearc blocks (Pezzopane and Weldon, 1993; Wells and others, 1998; Wells and Simpson, 2001). As a result, the southern portion of the forearc, the Oregon Coast block, is impinging on western Washington and northwest Oregon at a rate of about 8 to 12 mm/yr causing crustal shortening in northwest Oregon and western Washington (Wells and others, 1998; Wells and Simpson, 2001; Mazzotti and others, 2002).

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The combined effect of margin-normal subduction and margin-parallel shortening produces complex and diverse deformation within the northern edge of the Cascadia forearc and triggers large (greater than magnitude [M] 6), damaging earthquakes from three seismogenic source zones:

The locked zone of the CSZ fault interface, which produces great mega-thrust earthquakes;

The deep intraslab portion of the CSZ (i.e., the subducted portion of the Juan de Fuca Plate), the source of Wadati-Benioff zone earthquakes; and

The overriding North American Plate, where shallow crustal faults rupture.

All three sources potentially produce earthquakes that impact the ground motion hazards at the project site. Offshore, elastic release of strain accumulated in the locked plate interface of the CSZ produces great megathrust earthquakes (greater than M 8.0) occurring at irregular intervals that span from about 100 to more than 1,200 years, with an average recurrence interval of about 300 to 500 years (Atwater and Hemphill-Haley, 1997; Clague, 1997; Goldfinger and others, 2003 and 2012); the most recent rupture occurred in A.D. 1700 (Satake and others, 1996; Atwater and Hemphill-Haley, 1997; Clague, 1997; Yamaguchi and others, 1997; Goldfinger and others, 2003 and 2012).

Onshore, migration and rotation of tectonic blocks produce deformation along shallow faults within the upper part of the crust. At depth, rupture within the subducting slab, referred to as the intraslab, has produced some of the largest recorded earthquakes (M 6.5 to

7) to strike the Pacific Northwest, the northern California Coast, and Western Washington.

However, over the past century, intraslab earthquakes have been markedly infrequent in Oregon. The following sections briefly describe the location, characteristics, and seismicity of each of the sources.

2.1.1 Cascadia Subduction Zone: Mega-Thrust Interface Source

CSZ mega-thrust earthquakes originate along the interface between the subducting oceanic plates and the North American plate. Because of the significant uncertainty of the landward extent of a potential rupture surface, estimates of the closest distance between the project and potential rupture surface range from about 65 to 140 horizontal miles. Focal depths for mega-thrust earthquakes are commonly on the order of about 15 to 25 miles. Rupture of the interface could result in earthquakes with moment magnitudes (Mw) on the order of 8.5 to over 9.0, with strong shaking that lasts for several minutes. No large earthquakes have occurred in this zone during historic times (the last 170 years). However, geologic evidence suggests that coastal estuaries have experienced rapid subsidence at various times within the last 2,000 years (Atwater, 1987; Atwater and Hemphill-Haley, 1997) as a result of tectonic movement associated with mega-thrust earthquakes on the CSZ. It appears that

112510 July 2024 ruptures of this zone have occurred at irregular intervals that span from about 100 to more than 1,200 years, with an average recurrence interval of about 300 to 500 years (Atwater and Hemphill-Haley, 1997). Based on historical tsunami records in Japan (Satake and others, 1996) the most recent interplate event on the CSZ was a Mw 9 event on January 26, 1700.

2.1.2 Cascadia Subduction Zone: Intraslab Source

CSZ intraslab earthquakes originate from within the subducting oceanic plates as a result of down-dip tensional forces and bending caused by mineralogical and density changes in the plates at depth. These earthquakes typically occur 28 to 37 miles beneath the surface. The nearest seismogenic intraslab portion of the Juan de Fuca plate is approximately 30 to 60 miles below the Portland area. Ludwin and others (1991) estimate that the maximum Mw from this source zone would be about 7.5. Ground shaking produced by intraplate earthquakes would be less intense and less prolonged in the Portland area than ground motions generated by large subduction zone interface earthquake events. Historic seismicity from this source zone includes the 1949 Mw 6.7 Olympia earthquake, the 1965 Mw 6.7 earthquake between Tacoma and Seattle, and the 2001 Mw 6.8 Nisqually earthquake. While intraslab events have occurred frequently in the Puget Sound area, they are historically rare in Oregon.

2.1.3 Shallow Crustal Source

Shallow crustal earthquakes within the North American Plate have historically occurred in a diffuse pattern within the Pacific Northwest, typically within the upper 4 to 19 miles of the continental crust. Mabey and others (1993) concluded from their analysis of local geologic features that a crustal earthquake of up to Mw 6.5 could occur virtually anywhere in the Portland area. Based on their fault model, Wong and others (2000) determined that an earthquake of up to Mw 6.8 is possible on the Portland Hills Fault, which is mapped within about 2.4 miles of the project site. The largest known crustal earthquake in the Pacific Northwest is the 1872 North Cascades earthquake at approximate Mw 6.5 to 7.0. Other examples include the 1993 Mw 5.6 Scotts Mill earthquake and the 1993 Mw 6.0 Klamath Falls earthquake.

Faults and associated folds which demonstrate geological evidence of co-seismic surface deformation during the Quaternary time period in Oregon and Washington have been located and characterized by the United States Geological Survey (USGS). The USGS provides approximate fault locations and a detailed summary of available fault information in the USGS Quaternary Fault and Fold Database (2014). The database defines four categories of faults, Class A through D, based on evidence of tectonic movement known or presumed to be associated with large earthquakes during Quaternary time (within the last

2.6 million years). For Class A faults, geologic evidence demonstrates that a tectonic fault

112510 July 2024 exists and that it has likely been active within the Quaternary period. For Class B faults, there is equivocal geologic evidence of Quaternary tectonic deformation, or the fault may not extend deep enough to be considered a source of significant earthquakes. Class C and D faults lack convincing geologic evidence of Quaternary tectonic deformation or have been studied carefully enough to determine that they are not likely to generate significant earthquakes.

According to the USGS Quaternary Fault and Fold Database (USGS, 2024), there are 6 Class A features within approximately 15 miles of the project site. Their names, general locations relative to the site, and the time since their most recent deformation are summarized in Exhibit 2-1. The CSZ itself is approximately 135 miles west of the project site, with an average slip rate of approximately 40 millimeters (1.5 inches) per year.

Exhibit 2-1: USGS Class A Faults Within an Approximate 15-Mile Radius of the Project Site

Fault Name USGS Fault

Number Approximate

Length

Approximate Distance and Direction from Project Site1

Slip Rate Category2

Time Since Last Deformation3

Lacamas Lake Fault 880 14.9 miles 4.3 miles W < 0.2 mm/yr < 750 ka

East Bank Fault 876 18.0 miles 9.4 miles SW < 0.2 mm/yr < 750 ka

Portland Hills Fault 877 30.4 miles 10.3 miles SW < 0.2 mm/yr < 1.6 Ma

Oatfield Fault 875 18.0 miles 12.7 miles SW < 0.2 mm/yr < 1.6 Ma

Grant Butte Fault 878 6.2 miles 13.5 miles SE < 0.2 mm/yr < 750 ka

Damascus-Tickle Creek Fault Zone 879 9.9 miles 14.3 miles SE < 0.2 mm/yr < 750 ka

NOTES:

Approximate distance between the project site and nearest extent of fault mapped at the ground surface.

mm = millimeters; yr = year.

Ma = “Mega-annum” or million years ago; ka = “Kilo-annum” or one thousand years ago.

3 SUBSURFACE EXPLORATION PROGRAM

Shannon & Wilson explored subsurface conditions at the project site with two geotechnical borings, designated B-1 and B-2, and two test pits, designated TP-1 and TP-2. Borings B-1 and B-2 were completed to a depth of 71.5 feet, between February 15 and February 16, 2024.

Test pits TP-1 and TP-2 were completed to depths of 6.6 feet and 3.0 feet, respectively, on February 20, 2024. Infiltration testing was completed in TP-1 at a depth of 3.0 feet. To support pavement design, two dynamic cone penetration (DCP) tests, designated DCP-1, and DCP-2, were completed on February 20, 2024.

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Approximate locations of the explorations are shown on the Site and Exploration Plan, Figure 2. Details of drilling, sampling procedures, and our log of the materials encountered are presented in Appendix A, Field Explorations.

4 LABORATORY TESTING

The samples we obtained during field explorations were transported to the Shannon & Wilson laboratory for further examination. Representative samples were selected for a suite of laboratory tests. The testing program included moisture content analyses, particle size analyses, and corrosivity testing. Testing was performed by Shannon & Wilson in accordance with applicable ASTM International standards or subcontracted to an outside laboratory. Results of the laboratory tests and brief descriptions of the test procedures are presented in Appendix B, Laboratory Testing.

4.1 Corrosivity Testing Results

Analytical testing was performed on soil samples collected from the subsurface exploration to evaluate the corrosivity potential of the soil at the project site. The results of the corrosivity testing were graded for potentially corrosive environments to ductile-iron pipes using the 10-point evaluation procedure (AWWA, 2009). The evaluation procedure is based on information from soil resistivity, pH, oxidation-reduction potential, sulfides, and moisture. Based on the 10-point evaluation procedure, the analyzed soils are not considered corrosive to ductile-iron pipes.

The Naval Facilities Engineering Command Design Manual 7.02, Soils and Foundation Design Manuals (NAVFAC, 1986) indicates that as it pertains to foundations, if analysis indicates the sulfate concentration to be more than 0.5 percent in soil, or more than 1200 ppm in groundwater the use of a sulfate resistant cement such as Type V Portland cement should be considered. Measured sulfate concentrations were less than 0.5 percent in the soil and therefore below the criteria for using sulfate resistant cement.

5 SUBSURFACE CONDITIONS

5.1 Geotechnical Units

Shannon & Wilson grouped the materials encountered in our field explorations into a single geotechnical unit, as described below. The soil classifications are presented on the boring log. The materials encountered are described on the boring logs in Appendix A. A partial summary of the geotechnical unit is as follows:

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Catastrophic Flood Deposits: Lean Clay with Sand (CL); Sandy Silty to Silt with Sand (ML); Sandy Silt to Silty Sand (ML/SM); Silty Sand (SM); Poorly Graded Sand with Silt to Silty Sand (SP-SM/SM); Poorly Graded Sand with Silt (SP-SM); and Poorly Graded Sand to Poorly Graded Sand with Silt (SP/SP-SM).

The geotechnical unit was grouped based on the engineering properties, geologic origin, and distribution in the subsurface. Contacts between the units may be more gradational than shown in the boring logs. The Standard Penetration Test (SPT) N-values are shown on the boring logs as recorded in the field (uncorrected).

5.2 Groundwater

Shallow perched, groundwater was observed seeping through the side walls of the test pits and ponding at the base of the excavation. Groundwater was not noted on the borings B-1 or B-2 which were advanced using mud rotary drilling techniques. Mud-rotary drilling techniques can make the depth to groundwater difficult and/or not possible to discern during drilling due to the introduction of drilling fluids into the borehole to flush the drill cuttings to the surface. A vibrating wire piezometer (VWP) was installed in boring B-1 to allow for ongoing groundwater level measurements. A summary of groundwater measurements is provided in Exhibit 5-1.

Exhibit 5-1: Depth to Groundwater Summary

Exploration (Method) Date Depth of Water (ft)

TP-1 2/20/2024 3.6a

TP-2 2/20/2024 2.0a

B-1 3/6/2024 52.4b

B-1 3/26/2024 51.3b

B-1 6/17/2024 51.4

NOTES:

1. Measurement from visual observation of seepage in test pit at time of exploration was conducted.

2. Measurement from vibrating wire piezometer (VWP)

Groundwater levels should be expected to vary with changes in topography and precipitation. Generally, groundwater highs occur at the end of the wet season in late spring or early summer, and groundwater lows occur towards the end of the dry season in the early to mid-fall.

Based on the observations of groundwater in the test pit explorations and measurements collected from VWP, we anticipate a shallow perched layer of groundwater is present in the upper 5 feet of the subsurface at the site.

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5.3 Seismic Design Ground Motions

Our seismic hazard evaluation was performed in accordance with the criteria in the 2021 International Building Code (IBC). The 2021 IBC refers to ASCE's Minimum Design Loads for Buildings and Other Structures, 2016 Edition (ASCE 7-16) Chapter 20 for determination of Site Class. As discussed in Section 6.2, potentially liquefiable soils were encountered in the subsurface exploration program. Based on the potentially liquefiable soils encountered in our subsurface exploration, the site class assigned to the site is Site Class F.

As permitted by Chapter 20 of ASCE-16, site response analysis is not required to determine spectral accelerations for sites with potentially liquefiable soils (Site Class F) when the structure has a fundamental period of vibration equal to or less than 0.5 seconds. Provided the structure has a fundamental period of vibration equal to or less than 0.5 seconds, the site classification can be determined using ASCE 7-16 Table 20.3-1 and the site coefficients, Fa and Fv in ASCE 7-16 Tables 11.4-1 and 11.4-2. Our seismic analyses are based on the structure having a fundamental period of vibration of less than 0.5 seconds and a site response analysis is not required. Based on the Standard Penetration Test (SPT) N-values from the boring, the non-liquefied seismic site class for the site is Site Class D.

Based on Ss and S1 values mapped at the site by ASCE 7-16, Section 11.4.8 states a site response analysis is required for the site; however, we note some exceptions may be valid.

As permitted by Exception 1 in Section 11.4.8 of ASCE 7-16 Supplement 3, site response analysis is not required to determine spectral accelerations for a seismic Site Class D when the value of SM1, as determined by ASCE 7-16 Equation 11.4-2, is increased by 50 percent for all applications of SM1. Please note that SD1, as determined by ASCE 7-16 Equation 11.4-4, will then be increased by 50 percent for all applications in ASCE 7-16.

We have provided seismic design ground motion parameters that are based on Exception 1 in Supplement 3 being valid ( and that a site response analysis is not required. The provided USGS Code-Based MCE and Design Seismic Parameters for Site Class D are presented in Exhibit 5-2.

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Exhibit 5-2: USGS Code-Based MCE and Design Seismic Parameters

Seismic Parameter Value

Site Class D

Mapped MCE Peak Ground Acceleration, PGA 0.363 g

PGA Site Coefficient, FPGA 1.237

Peak Ground Acceleration Corrected for Site Effects, PGAM 0.449 g

Mapped Short Period Spectral Acceleration, Ss 0.806 g

Mapped 1-Second Period Spectral Acceleration, S1 0.367 g

Short Period Site Coefficient, Fa 1.178

1-Second Period Site Coefficient, Fv 1.933

Short Period Design Spectral Acceleration, SDS 0.633 g

1-Second Period Design Spectral Acceleration, SD1 0.709 g*

NOTES:

Due to the potentially liquefiable soils at the site, the seismic site class per ASCE 7-16 is F; however, the seismic parameters to determine base shear forces can be determined using the non-liquified seismic site class (Site Class D) provided the structure has a fundamental period of vibration less than 0.5 seconds per Chapter 20 of ASCE 7-16. The seismic parameters provided in Exhibit 5-2 are based on Site Class D.

g = acceleration due to gravity

* See Section 11.4.8 of ASCE 7-16 Supplement 3. Value of SD1 in Exhibit 5-2 has been increased by 50 percent as required by ASCE 7- 16 Supplement 3.

5.4 Liquefaction

5.4.1 General

Soil liquefaction is a phenomenon in which excess pore water pressure of loose to medium dense, saturated, granular soils increases during ground shaking to a level near the initial effective stress. The increased excess pore pressure results in a reduction of soil shear strength. The effects of liquefaction typically include lateral spreading, slope instability, and ground settlement.

Soil behavior under seismic loading is the primary factor in determining the susceptibility of soil to liquefaction. Important factors in evaluating soil behavior are relative density, the fines content (percent of soil by weight smaller than 0.075 millimeters, passing the No. 200 sieve), and the plasticity characteristics of the fines. Relative density can be estimated from SPT N-values collected from the borings for this project.

Settlement may occur in cohesionless soil that undergoes liquefaction and pore pressure development during ground shaking. The settlement is related to densification and rearrangement of particles during ground shaking, as well as volume change as the excess pore pressure dissipates after ground shaking. Seismic ground settlement may not occur

112510 July 2024 uniformly over an area, and differential settlement may impact existing or proposed structures supported by liquefied soil.

5.4.2 Liquefaction Analysis and Liquefaction-Induced Settlement

We evaluated liquefaction potential of the upper 70 feet of the site soils based on the SPT N-values collected from borings B-1 and B-2, results of laboratory testing, and groundwater measurements collected from the VWP installed in boring B-1. These collective data were evaluated according to the methods and procedures recommended by Boulanger and Idriss (2014). In addition, our analysis was performed using a peak ground acceleration (PGAM) of 0.449g and a mean Moment Magnitude (Mw) of 7.8, which is based on the total seismic source contribution to the site. The mean Mw was determined by using USGS’s online Unified Hazard Tool for a 2,475-year return period.

Liquefaction-induced settlement magnitudes were calculated using correlations developed by Tokimatsu and Seed (1987) and Ishihara and Yoshimine (1992). Results of our liquefaction-induced settlement analysis indicate that total liquefaction-induced settlement will be up to 1.2 inches with the settlement occurring within a potentially liquefiable zone of soil between depths of 50 and 60 feet bgs. Exhibit 5-3 presents a summary of the liquefaction-induced volumetric strain calculated at boring locations B-1 and B-2.

Exhibit 5-3: Summary of Liquefaction-Induced Settlement at Depth

Boring Location

Approximate Depths of Liquefaction Susceptible Soils

(feet)

Approximate Total Liquefaction- Induced Settlement

(inch)

B-1 52 to 58 0.8 to 1.2

B-2 65 to 70 Up to 0.3

Ishihara and others (2016) provided an update to the procedures included in Ishihara and Yoshimine (1992) for estimating settlement triggered by liquefaction at the ground surface recommended procedure. In their recommendations, they state that volumetric strain estimated in loose liquefiable layers below a depth of 15 meters (49.2 feet) should not be counted in settlements manifesting at the ground surface. As the depth of the static groundwater table exceeds this depth, liquefaction-induced settlement that manifests at the ground surface is anticipated to be minimal.

5.5 Lateral Spreading

Based on the depth to groundwater and distance from existing free faces at the project site, we consider the risk of lateral spreading impacting the new facility to be low.

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5.6 Fault Surface Rupture

As presented in Exhibit 2-1, the closest fault to the site is mapped approximately 4.3 miles to the west. Consequently, it is our opinion that the risk of surface fault rupture at the site is low during a design life of 100 years or less.

6 GEOTECHNICAL ENGINEERING

RECOMMENDATIONS

6.1 General

The key geotechnical design considerations for the site include settlement due to long-term loading and the seismic considerations. As discussed in Section 5.4.2, liquefaction-induced ground settlement at the ground surface is expected to be minimal given the distance to the groundwater table (and potentially liquefiable layers) beneath the ground surface.

6.2 Foundation Design – Tanks and Vessels

The proposed six Model 12-40 GAC's are 12-feet in diameter and 26 feet and 9 inches in height. Each GAC vessel will have a system operating weight of 391,000 pounds. We understand all six GAC vessels will be supported on a structural pad. Based on discussion with the structural engineer, we understand the average contact pressure at the base of the structural pad will be less than 3,000 psf.

We do not have information on the configuration of the 80,000-gallon backwash tank;

however, we anticipate it will by cylindrical and will likely have a height and diameter on the order of 40 feet and 18 feet, respectively. Based on discussion with the structural engineer, we understand the backwash tank will have an average contact pressure less than 3,000 psf at its base.

Given the relatively high areal pressurebeneath the tanks and vessels, we recommend installing a minimum 36-inch-thick reinforced crushed rock mat beneath the tank and vessel foundations. Details regarding the reinforced crushed rock mat is presented in Section 7.4.

For foundations bearing on the reinforced crushed rock mat section, a net allowable bearing pressure of 3,000 psf may be used for design. We further recommend evaluating the tank vessel foundations with a subgrade reaction modulus of 150 pci.

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6.2.1 Tank and Vessel Settlement

Based on the above loading configurations and recommended implementation of reinforced crushed rock mats, we anticipate total settlement beneath the vessels and backwash tank will be up to 1 inch. Differential settlements across the vessels and tanks are expected to be up to 50 percent of the estimated total settlement.

6.2.2 Tank and Vessel Lateral Resistance

The nominal lateral sliding resistance at the base of the tank and vessel foundations can calculated by multiplying the vertical load by a coefficient of 0.4 for provided the foundation is underlain by the crushed rock mat described above in Section 6.2 and below in Section 7.4

6.3 Foundation Design – Restrooms/Storage Building

We understand the restroom/storage building will be consructed at-grade and will consist of cement modular unit (CMU) construction. According to ASCE 7-16 Table 12.13-3, single-story structures with concrete or masonry wall systems have differential settlement value limits of 1.2 inches over a distance of 50 feet for risk category IV structures. Wall loads for the restroom building are anticipated to be 1 kip per lineal foot or less.

The bottom of the foundation should be at least 18 inches below the lowest adjacent grade for frost protection. Based on a bottom of footing depth of 18 inches, the wall footings of the restroom/storage building are anticipated to be founded on loose to medium dense silty sand of the Catastrophic Flood Deposits unit. Accordingly, we recommend constructing a minimum 6-inch-thick crushed rock pad beneath the footings to help provide a more uniform surface and protect the subgrade from becoming disturbed during placement of rebar and formwork. The crushed rock pad should be compacted to a well-keyed, firm, and unyielding condition.

While not encountered in our explorations, if undocumented fill is encountered during construction, the undocumented fill should be removed from the foundation subgrade and be replaced with crushed rock. The crushed rock should extend 6 inches horizontally outward from the footing edge for every 1 foot of thickness (i.e., 0.5H:1V outward projection from edge of foundation). The crushed rock should be compacted to not less than 93percent of the maximum dry density, as determined by ASTM D1557.

6.3.1 Allowable Bearing Capacity – Restroom/Storage Building

We recommend an allowable bearing capacity of 2,000 pounds per square foot (psf) be used to design the building foundations provided the minimum 6-inch-thick crushed rock pad is

112510 July 2024 incorporated into the design. This allowable bearing capacity can be increased by up to one-third for short-term loads resulting from wind or seismic forces. We recommend the wall footings have a minimum width of 18 inches. If the wall foundation is considered to be flexible, a subgrade reaction modulus of 125 pounds per cubic inch (pci) may also be used in design.

6.3.2 Building Foundation Settlement

Based on the above allowable bearing capacity of 2,000 psf and with the understanding that the wall loads will be less than 1 kips per lineal foot, we anticipate total settlement beneath the wall footings will be less than 1 inch. Differential settlement is anticipated to be less than 50 percent of the total settlement which will be less that the threshold provided in ASCE 7-16 Table 12.13-3 for single-story structures with concrete or masonry wall systems.

6.3.3 Foundation Lateral Resistance – Restroom/Storage Building

The soil resistance available to withstand lateral foundation loads is a function of the frictional resistance, which can develop on the base of the foundation, and the partial soil passive resistance, which is assumed to be about 50 percent of the full soil passive resistance. The nominal lateral sliding resistance at the base of the footing can calculated by multiplying the vertical load by a coefficient of 0.4 for footings underlain by a minimum of 6 inches of crushed rock.

A nominal partial passive fluid pressure of 200 pcf may be also used for design. Passive resistance should be ignored in the upper 12 inches as construction and post-construction activities often disturb this upper material and it is susceptible to frost penetration.

6.4 Lateral Earth Pressures on Buried Structures and Retaining Walls

6.4.1 Static and Seismic Lateral Earth Pressures

We recommend that the embedded walls of below grade vaults and retaining walls less than 4 feet tall be evaluated as non-yielding walls by applying a static at-rest equivalent fluid earth pressure equal to 56H (where H is the height of the wall retaining soil) which is based on level retained material.

We understand a interlocking block retaining wall will support a backslope with inclinations up to 3.3H:1V. For wall locations with a maximum backslope of 3H:1V, we recommend evaluating the retaining wall with a static at-rest equivalent fluid earth pressure of 73H (where H is the height of the retained soil). These static at-rest earth pressures should be evaluated as a triangular distribution.

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Section 1807.2.2 of the 2021 International Building Code (currently adopted by City of Vancouver at the time of this report) requires retaining walls supporting retained material up to 6 feet in height be designed to account for seismic lateral earth pressure. Evaluation of seismic loading should be performed by applying an additional uniform seismic earth pressure of 25H on portions of the wall retaining soil. The additional seismic earth pressure should be evaluated as a uniform distributed pressure increase. This seismic earth pressure should be applied with the static at-rest pressure value. Additional surcharge pressures may be appropriate for evaluation if dead of live loads, such as traffic loads, are anticipated to be present.

6.4.2 Embedded Wall Drainage

Hydrostatic pressures are not included in the equivalent fluid pressures above as the long-term static groundwater is anticipated to be below the base of the embedded portions of the structures; however, we note that perched groundwater conditions can develop within a few feet of the ground surface at the site. To prevent the buildup of hydrostatic pressures on the vaults from perched water, we recommend that any vault structure with a height greater than 4 feet be backfilled with a minimum 12-inch-wide zone of free-draining crushed rock material (i.e., 1 1/2-inch minus with less than 3 percent passing the No 200 sieve). To protect this 12-inch-wide zone of crushed rock, we recommend placing a non-woven geotextile (such as Mirafi 140 N) along the contacts of the crushed rock zone with the adjacent and overlying soil. A drain should be included at the base of the crushed rock zone to reroute collected water to a suitable discharge point.

6.4.3 Surcharge Loading

Surcharge loads, if applicable, should be added to static and seismic lateral earth pressures.

If vehicle traffic will be operating near the walls, then typically a 250 psf vertical surcharge load is applied as a lateral uniform pressure equal to 118 psf. If construction traffic surcharge pressures in excess of 250 psf will be applied at the ground surface adjacent to the buried building wall during or after construction, we recommend that those surcharge pressures be modeled as a uniform rectangular pressure acting on the face of the non-yielding wall as 0.47q, where q is the surcharge load as shown in Figure 3. Other surcharge loading conditions should be evaluated, as applicable, with the use of the general equations and diagrams shown on Figure 4.

6.4.4 Lateral Resistance

Lateral resistance of buried vault structures can be obtained through partial passive pressure developed along the buried portions of the structures. Provided the structures are backfilled with crushed rock compacted to at least 95 percent of the maximum dry density

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(ASTM D1557), passive resistance on buried foundations can be calculated based on granular backfill material with a friction angle of 34 degrees, a unit weight of 120 pounds per cubic foot (pcf), and the equivalent fluid lateral pressures in Exhibit 6-1. Note, the full passive value is based on lateral movement equal to up to 6 percent of the wall height occurring before the full passive resistance is reached.

Exhibit 6-1: Passive Equivalent Fluid Pressures

Coefficient Full Passive Value

(psf) Partial Passive Value

(psf)

Passive 350H 175H H = embedded height of structure (feet)

For estimating the lateral resistance contribution from the partial passive pressure, we recommend ignoring the upper 1-foot of adjacent backfill or soil due to potential disturbance resulting from post-construction activities and frost penetration.

6.5 Pavement Design

We understand that the proposed pavement will be subjected to traffic loading from passenger vehicles and occasional chemical trucks during normal operations. The pavement design was performed in accordance with the recommended procedures and guidelines in the 1993 AASHTO Guide for Design of Pavement Structures (AASHTO, 1993).

The pavement section thicknesses recommended below can accommodate up to 100,000 Equivalent Single Axle Loads ESALs over a 20-year design life, which correlates to average annual daily traffic of approximately 3 to 7 passes from a 2-axle freight truck. The traffic conditions were discussed with the project team and were confirmed to be a suitable minimum loading condition for the proposed pavement. However, the project team indicated that a more conservative estimate would be 120,000 ESALs over a 20-year design life. The increase from 100,000 to 120,000 ESALs is a relatively small increase in loading for pavement and does not significantly increase the pavement thickness. Although, for comparison we have presented pavement thicknesses for both loading conditions presented in Exhibit 6-2 below.

Based on the results from the DCP testing, we estimated a resilient modulus (Mr) based on an empirical correlation to the DCP penetration. The Mr value provides an indication of the underlying subgrade strength. The Mr value correlated from the DCP testing is estimated to be an average of 3,000 psi. Subgrades with Mr values less than 5,000 psi are generally considered to be poor subgrade conditions for pavement. Other inputs for the AASHTO pavement analysis (i.e., layer coefficients, standard deviation, reliability) are based on the recommended values presented in the design guide.

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Exhibit 6-2: Recommended Pavement Thicknesses

Loading Condition

Design Life (years)

AC Thickness (inches)

Base Aggregate Thickness (inches) Comment

100,000 ESALs 20

4.0 12.0 Minimum Thickness

5.0 9 Alternate Thickness

120,000 ESALs 20

4.0 13.0 Minimum Thickness

5.0 10.0 Alternate Thickness

Due to the relatively low subgrade strength measured in our DCPs, we would recommend inclusion of a woven geotextile such as Mirafi RS380i or RS580i or equivalent be included between the subgrade and the base aggregate. A woven geotextile with properties similar to Mirafi RS380i is capable of providing both separation and some reinforcement of the pavement section and will improve the long-term performance of the pavement.

7 CONSTRUCTION CONSIDERATIONS

7.1 Earthwork

Site preparation will include clearing, grubbing, and stripping and excavation during subgrade preparation. Organic material and topsoil should be stripped and removed from all proposed building and pavement areas. Based on our test pit explorations, we anticipate a stripping depth of approximately 6 to 7 inches. Greater depths may be necessary to remove localized zones of organic material. Stripped material should be transported off site for disposal or used as fill in landscaping areas.

We recommend that the primary root systems for trees and other vegetation be completely removed. Trees designated for preservation should be clearly marked prior to site stripping, clearing, and grubbing. Trees and their root balls should be grubbed to the depth of the roots, which would exceed 3 feet bgs. Depending on the methods used to remove the root balls, considerable disturbance of the subgrade could occur during site clearing and grubbing. We recommend that soil disturbed during clearing and grubbing operations be removed and replaced with compacted crushed rock structural fill.

7.2 Subgrade Preparation

As recommended in Section 6.2, a 36-inch-thick reinforced crushed rock mat is recommended to be installed beneath the tank and vessel structural slab foundations a reinforced crushed rock. In Section 6.3, a minimum 6-inch-thick crushed rock pad is recommended to be installed beneath the restroom/storage facility footings. All subgrades beneath these recommended crushed rock sections; as well as the pavement subgrades

112510 July 2024 should be neatly trimmed and carefully prepared. Any deleterious, loose, or softened material should be removed from the subgrade and replaced with crushed rock and any loose subgrade compacted prior to placing crushed rock. All crushed rock beneath foundations and pavement structures should be compacted to not less than 95 percent of the maximum dry density as determined by ASTM D1557.

Prior to the placement of crushed rock or the construction of foundations, structural slabs, or pavements, we recommend proof rolling the subgrade with a 10-to-12-ton roller operating in the static mode to identify any areas of excessive yielding. The proof rolling should be observed by the Geotechnical Engineer of Record or their representative who will evaluate the subgrade. If areas of excessive yielding are identified, the material in the yielding subgrade should be removed and replaced with compacted granular structural fill.

Areas that cannot be accessed with proof rolling compaction equipment should be evaluated by probing by the Geotechnical Engineer of Record or their representative. In addition, probing should be used to evaluate subgrade in pavement areas if the subgrade consists of primarily fine-grained soil that is wet and sensitive. Soft or weak spots should be over-excavated and replaced with imported compacted granular material. We recommend a contingency be placed in the budget for further over-excavation and stabilization of the subgrade as necessary.

We recommend that the footing excavations be observed by the Geotechnical Engineer of Record or their representative prior to placing steel and concrete to evaluate the suitability of the exposed subgrade, and to verify that the recommendations of this report have been followed and that conditions encountered are as anticipated. All deleterious, soft, or unsuitable materials observed by the Geotechnical Engineer should be removed and replaced with crushed rock compacted to not less than 95 percent of the maximum dry density determined by ASTM D1557.

7.3 Wet Weather Construction

Excavation and construction operations may expose the on-site soils that are sensitive to inclement weather conditions. The stability of exposed soils may rapidly deteriorate due to a change in moisture content (i.e., wetting or drying) or the action of heavy or repeated construction traffic. Accordingly, excavations should be adequately protected from the elements and from the action of repetitive or heavy construction loadings. Exposed subgrade should be covered with crushed rock as soon as practical to prevent water from softening the subgrade resulting in the need for additional over-excavation. Exposed subgrade should not be left open overnight during the rainy season which typically extends from mid-October to June. If water accumulates in the excavation for any reason, prior to

112510 July 2024 placing crushed rock, the water should be sump pumped out with an engineered sump system to protect from removal of soils fines during pumping, and subgrade surface reobserved by the Geotechnical Engineer of Record.

7.4 Reinforced Crushed Rock Mat

The subgrade beneath the reinforced crushed rock mat should be prepared as described in Sections 7.1 and 7.2 of this report. A detail showing the reinforced crushed rock mat is presented in Figure 5.

After the subgrade has been approved by the Owner’s Representative, we recommend placing a layer of non-woven geotextile conforming to the properties provided in Table 3, Section 9-33.2(1) of the Washington Department of Transportation Standard Specifications, such as Mirafi 180N. The non-woven geotextile layer should be immediately overlain by a geogrid such as Tensar InterAx NX750 or NX850 Geogrid or an approved alternative. An alternative to having separate geotextile layers (i.e., separate non-woven and geogrid materials) overlying the subgrade is to use a geotextile with both separation and reinforcing properties such as Mirafi RS580i. All geotextile and geogrid layers should be placed over the entire surface of the subgrade and joints overlapped and/or tied (in the case of geogrid) in accordance with the manufacturer’s recommendations. The geogrid described above should not be placed along uneven surfaces or along localized low areas created by additional over-excavations. As shown on Figure 5, the geogrid should be placed as a horizontal layer.

The backfill placed over and between the geogrid layers should be structural fill material as described in Section 7.5 of this report. Within the crushed rock fill, we recommend placing a layer of geogrid at 12-inch intervals during backfill placement, as shown in Figure 5.

7.5 Structural Fill Placement and Compaction

Unless otherwise specified or shown, we recommend…