[Supplemental]_100343_Final_Geotech_Report.pdf

REPORT OF GEOTECHNICAL ENGINEERING SERVICES

Century Boulevard Extension SE Century Boulevard Bridge over Rock Creek Hillsboro, Oregon

For Washington County Department of Land Use and Transportation c/o NV5 March 7, 2025

Project: WashCo-78-01

March 7, 2025

Washington County Department of Land Use and Transportation c/o NV5 9450 SW Commerce Circle, Suite 300 Wilsonville, OR 97070

Attention: Scott Zelenka, P.E.

Report of Geotechnical Engineering Services Century Boulevard Extension

SE Century Boulevard Bridge over Rock Creek Hillsboro, Oregon

Project: WashCo-78-01

NV5 is pleased to submit this geotechnical engineering services report for the proposed Century Boulevard Extension project in Hillsboro, Oregon. This report presents a description of our services and site conditions; a summary of field explorations and laboratory testing; and recommendations for design and construction of bridge foundations and pavement design parameters.

We appreciate the opportunity to be of service to Washington County. Please contact us if you have questions regarding this report.

Sincerely, NV5

Hayden Gluck, P.E.

Associate Engineer

JDT:HDG

Attachments

One copy submitted

Document ID: WashCo-78-01-030725-geor.docx

© 2025 NV5. All rights reserved.

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EXECUTIVE SUMMARY

Based on information provided by Washington County and the results of our explorations, laboratory testing, and analyses, it is our opinion that the project is feasible. The primary geotechnical considerations for the project are summarized as follows:

We recommend that the bridge be supported on driven 16-inch-diameter pipe piles that extend into the stiff to very stiff silt and clay.

MSE walls can be constructed to support bridge approach fills, and CIP walls can support the detention pond fill east of the south abutment wall. The estimated total settlement of the bridge approach walls will range from 3.5 to 5 inches and should essentially be complete within four months after completion of wall construction.

Based on the cyclic direct shear testing, the soil at the site will not liquefy during a design-level seismic event. However, a 15 percent reduction in shear strength will occur during the design seismic event. Based on our stability analyses and considering the anticipated shear strength reduction, lateral spreading and slope stability are not a design consideration at the site.

The near-surface soil at the site contains fine-grained soil that is easily disturbed during wet weather. The contractor should protect the subgrade areas from disturbance during construction.

WashCo-78-01:030725

TABLE OF CONTENTS

ACRONYMS AND ABBREVIATIONS

1.0 INTRODUCTION 1

2.0 PURPOSE AND SCOPE 1

3.0 SITE CONDITIONS 2

3.1 Geologic Setting 2

3.2 Surface Conditions 3

3.3 Subsurface Conditions 3

4.0 CONCLUSIONS 5

5.0 DESIGN 5

5.1 Permanent Slopes 5

5.2 Seismic Considerations 5

5.3 Bridge Foundation Support 8

5.4 Signal Pole Foundation Design 14

5.5 Bridge Approach and Detention Pond Fills 15

5.6 Retaining Structures 16

5.7 Drainage 21

6.0 CONSTRUCTION 21

6.1 Erosion Control 21

6.2 Site Preparation 21

6.3 Wet Weather Considerations 22

6.4 Excavation 23

6.5 Materials 24

7.0 OBSERVATION OF CONSTRUCTION 27

8.0 LIMITATIONS 27

REFERENCES 29

FIGURES

Vicinity Map Figure 1 Site Plan – South Figure 2 Site Plan – North Figure 3 Foundation Data Sheet Figure 4

TABLE OF CONTENTS

APPENDICES

Appendix A Field Explorations A-1 Laboratory Testing A-1 Exploration Key Table A-1 Soil Classification System Table A-2 Boring Logs Figures A-1 – A-9 Atterberg Limits Test Results Figure A-10 Consolidation Test Results Figure A-11 Grain-Size Test Results Figure A-12 Summary of Laboratory Data Figure A-13 SPT Hammer Calibration Appendix B 2002 Field Explorations B-1 Site Plan, Exploration Logs, Laboratory Testing Results Appendix C Driven Pile Axial Resistance C-1 Factored Axial Resistance – Bent 1 Figure C-1 Factored Axial Resistance – Bent 2 Figure C-2 Factored Axial Resistance – Bent 3 Figure C-3 Nominal Axial Resistance – Bent 1 Figure C-4 Nominal Axial Resistance – Bent 2 Figure C-5 Nominal Axial Resistance – Bent 3 Figure C-6 Appendix D GROUP Analyses D-1 Century Boulevard Bridge Foundation Loads Table D-1 Analyses Results Appendix E Slope Stability Analyses E-1 Analyses Results

ACRONYMS AND ABBREVIATIONS

AASHTO American Association of State Highway and Transportation Officials AOS apparent opening size ASTM American Society for Testing and Materials BGS below ground surface CIP cast-in-place CSZ Cascadia Subduction Zone CSZE Cascadia Subduction Zone Earthquake FOS factor of safety g gravitational acceleration (32.2 feet/second2) GDM Geotechnical Design Manual H:V horizontal to vertical km kilometers ksf kips per square foot LRFD Load and Resistance Factor Design MSE mechanically stabilized earth ODOT Oregon Department of Transportation OSHA Occupational Safety and Health Administration OSSC 2021 Oregon Standard Specifications for Construction pcf pounds per cubic foot pci pounds per cubic inch PDA Pile Driving Analyzer® PGA peak ground acceleration psf pounds per square foot psi pounds per square inch SPT standard penetration test WEAP wave equation analysis program

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1.0 INTRODUCTION

This report presents the results of NV5’s geotechnical engineering evaluation for the proposed Century Boulevard Extension project in Hillsboro, Oregon. The site relative to surrounding features is shown on Figure 1. The project layout is shown on Figure 2. Acronyms and abbreviations used herein are defined above, immediately following the Table of Contents.

Based on information provided by NV5, the project will consist of a new bridge over Rock Creek and new roadway alignment along SE Century Boulevard south of West Baseline Road to SE Borwick Street. The bridge will be a two-span structure with a total length of 400 feet. Approach embankments will be supported by MSE retaining walls up to 23 feet in height above finished grade, and a detention pond will be constructed on the east side of the south bridge approach that will include a CIP retaining wall with a maximum height of 12.7 feet above finished grade.

We understand that a mast arm signal pole with a drilled shaft foundation will be installed at the intersection of SE Century Boulevard and SE Lois Street.

2.0 PURPOSE AND SCOPE

The purpose of our geotechnical engineering services was to explore the subsurface conditions near the proposed improvements and develop geotechnical engineering recommendations for design and construction of the proposed bridge, signal pole, and retaining walls. Our scope included the following:

Coordinated and managed the field explorations, including locating utilities, access preparation, and scheduling contractors and NV5 staff.

Prepared traffic control plans and obtained right-of-way permits from Washington County

(County).

Drilled eight borings to depths between 16.5 and 121.5 feet BGS for use in design of the bridge foundations, retaining walls, and signal pole foundation.

Drilled one hand auger boring and collected one surface sample for hydrometer testing for scour analysis.

Maintained documentation of each exploration, visually classified the soil encountered, and collected soil samples as appropriate for the soil conditions encountered.

Conducted the following laboratory tests on soil samples from the explorations:

Ninety-five moisture content determinations in general accordance with ASTM D2216 Two dry density determinations in general accordance with ASTM D7263 Two grain-size analyses in general accordance with ASTM D422 Thirteen particle-size analyses in general accordance with ASTM D1140 Two consolidation tests in general accordance with ASTM D2435 Six Atterberg limit tests in general accordance with ASTM D4318 Four cyclic direct simple shear tests.

Provided geotechnical engineering construction recommendations for site preparation, structural fill compaction criteria, wet/dry weather earthwork procedures, cut and fill slopes, and suitability of on-site soil for use as structural fill.

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Provided recommendations for design and construction of bridge foundations, including nominal and factored driven pile resistances, downdrag loads, and seismic design parameters. Provided estimated settlements and soil parameters for lateral load design.

Provided the results of the lateral analysis of pile groups for the bridge foundations.

Provided this report presenting the results of our field explorations and geotechnical engineering recommendations.

3.0 SITE CONDITIONS

3.1 GEOLOGIC SETTING

3.1.1 General

The site is located in the Tualatin Basin of the Puget Sound-Willamette Valley physiographic province, a tectonically active lowland located along the convergent Cascadia margin (Orr and Orr, 1999). The Tualatin Basin is formed by a gentle syncline between the uplifted Coast Ranges to the west, the Chehalem Mountains to the south, and the Tualatin Mountains to the north and east. The Tualatin Mountains have been uplifted along northwesterly oriented faults, including the steeply dipping Portland Hills fault located along the eastern flank of the mountains.

The near-surface geologic unit mapped at the site is the fine-grained facies of the Quaternary flood deposits (Madin, 1990). The unit consists of unconsolidated silt, clay, and sand deposited by catastrophic floods associated with the sudden release of waters from glacial Lake Missoula during the late Pleistocene Age. Many dozens of these Missoula floods occurred between approximately 15,500 and 12,500 years ago (and perhaps during earlier glaciations). Flood waters several hundred feet deep swept out of the Columbia Gorge and over the lowlands of the Portland area. The thickness of the flood deposits in the site vicinity is approximately 50 to 60 feet.

Underlying the Quaternary flood deposits is a thick accumulation of moderately to poorly lithified mudstone and sandstone mapped as the Sandy River Mudstone equivalent (Madin, 1990).

These were deposited in a large delta where the ancient Columbia and Willamette rivers converged, probably during the middle Miocene to Pleistocene Age (approximately 6 million to 1 million years ago). The thickness of the unit in the site vicinity is approximately 250 feet.

Basement rocks underlying the Sandy River Mudstone equivalent in the site vicinity consist of the Miocene Columbia River Basalts, emplaced approximately 17 million to 6 million years ago in the Portland area (Madin, 1990). The Columbia River Basalts are exposed in the Tualatin Basin in the highlands surrounding the valley and in a group of mountains south of the site, which include Cooper Mountain and Bull Mountain. The Columbia River Basalts consist of thick flows of basalt erupted from fissures in eastern Oregon, Washington, and western Idaho that traveled down the ancient Columbia River Gorge to fill the lowland areas around Portland.

3.1.2 Seismicity

Three scenario earthquakes are possible within the local seismic setting. Two of the possible earthquake sources are associated with the CSZ, and the third event is a shallow, local crustal earthquake that could occur in the North American Plate. The three earthquake scenarios are discussed below.

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The CSZ is the region where the Juan de Fuca Plate is being subducted beneath the North American Plate. This subduction is occurring in the coastal region between Vancouver Island and northern California. Evidence has accumulated suggesting that this subduction zone has generated eight great earthquakes in the last 4,000 years, with the most recent event occurring approximately 300 years ago (Weaver and Shedlock, 1991). The fault trace is mapped approximately 50 to 120 km off the Oregon Coast.

Two types of subduction zone earthquakes are possible:

1. An interface event earthquake on the seismogenic part of the interface between the Juan de Fuca Plate and the North American Plate on the CSZ. This source is reportedly capable of generating earthquakes with a moment magnitude of between 8.5 and 9.0.

2. A deep intraplate earthquake on the seismogenic part of the subducting Juan de Fuca Plate. These events typically occur at depths of between 30 and 60 km. This source is capable of generating an event with a moment magnitude of up to 7.5.

A significant earthquake could occur on a local fault near the site within the design life of the bridge. Such an event would cause ground shaking at the site that could be more intense than the CSZ events, although the duration would be shorter. The most significant fault in the site vicinity is the Helvetia fault, which is located approximately 5 km north of the site. Other significant crustal faults are the Beaverton fault zone and the Oatfield fault, which range from approximately 10 to 11 km from the site.

3.2 SURFACE CONDITIONS

The proposed road alignment between West Baseline Road and SE Borwick Street consists of undeveloped land. The northern one-third of this alignment is covered with heavy tree and shrub vegetation. The southern two-thirds of this alignment is covered with grass vegetation. Rock Creek is located near the transition between the tree and grass vegetation areas. Rock Creek is presently approximately 20 feet wide, 2 to 3 feet deep, and contains relatively slowly flowing water. Elevations along the bridge alignment are approximately 163 feet at the north end, 169 feet at the south end, and 128 feet at the center at Rock Creek.

3.3 SUBSURFACE CONDITIONS

3.3.1 General

The subsurface conditions were explored by drilling seven borings (B-2 through B-6, B-8, and B-9) along the proposed road alignment between West Baseline Road and SE Borwick Street and one boring (B-7) at the intersection of SE Century Boulevard and SE Lois Street. Three of the borings (B-2 through B-4) were drilled at the bridge location to a depth of 121.5 feet BGS. A boring from a previous investigation for the Baseline Road improvements in 2002 was completed at the intersection of West Baseline Road and SE Century Boulevard to a depth of 121.5 feet BGS. Four of the borings (B-5, B-6, B-8, and B-9) were drilled at the south approach embankment to depths between 16.5 to 31.5 feet BGS. One hand auger boring was completed near Rock Creek to a depth of 7.5 feet BGS. Laboratory testing to determine physical and engineering properties of the soil at the site was completed on select samples from the explorations.

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The general subsurface profile consists of alluvial silt extending to a depth of 5 feet BGS underlain by catastrophic flood deposits of silt, clay, and sand that extend to depths between 45 and 60 feet BGS. The catastrophic flood deposits are underlain by deposits of Sandy River Mudstone equivalent, which consists of silt, clay, and sand that extend to the maximum depth explored of 121.5 feet BGS.

The approximate locations of the explorations are shown on Figures 2 and 3. Exploration logs and laboratory test results for our 2019 investigation are presented in Appendix A. An exploration log and laboratory test results for our 2002 investigation are presented in Appendix B.

3.3.2 Silt (Fill and Recent Alluvium)

Recent alluvium consisting of very soft to soft silt overlying medium stiff to very stiff silt and clay was encountered immediately below the ground surface. Laboratory testing of select samples indicates the recent alluvium deposits had in-place moisture contents between 33 and 34 percent at the time of our explorations. The silt generally has low to moderate plasticity. Very soft to soft silt and clay fill was encountered in boring B-3 below the ground surface extending to a depth of 5 feet BGS.

3.3.3 Silt, Clay, and Sand (Catastrophic Flood Deposits)

Silt, clay, and sand catastrophic flood deposits were encountered below the recent alluvium. The catastrophic flood deposits extend to depths between 45 and 60 feet BGS. The silt and clay are generally medium stiff to stiff with isolated zones that are soft. Laboratory testing of select samples indicates the silt and clay had in-place moisture contents between 27 and 41 percent at the time of our explorations. Sand layers were encountered within the catastrophic flood deposits in boring B-3 between depths of 21 and 22.5 feet BGS , from a depth of 20 feet BGS to the 21.5-foot BGS maximum exploration depth in boring B-6, and between depths of 16 and 21 feet BGS in boring B-7. The sand is loose to dense with SPT values ranging from 5 to 38 blows per foot.

3.3.4 Silt, Clay, and Sand (Sandy River Mudstone Equivalent)

We encountered Sandy River Mudstone equivalent deposits of silt, clay, and sand below the catastrophic flood deposits that extend to the maximum depth explored of 121.5 feet BGS. We interpret these soils to represent decomposed mudstone overlying the intact mudstone. The silt and clay are generally stiff to very stiff. Laboratory testing of select samples indicates the silt and clay had in-place moisture contents between 23 and 51 percent at the time of our explorations.

Medium dense to dense sand of the Sandy River Mudstone equivalent deposits was encountered from a depth of 120 feet BGS to the maximum depth explored of 121.5 feet in boring B-2, between depths of 80 and 85 feet BGS in boring B-3, and from a depth of 114 feet BGS to the maximum depth explored of 121.5 feet BGS in boring B-4. SPT values ranged from 23 to 31 blows per foot.

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3.3.5 Groundwater

The presence of groundwater was difficult to directly determine due to the mud rotary drilling techniques in the explorations. Based on moisture increases in the collected samples, we believe the groundwater ranges from approximately 5 to 15 feet BGS. We anticipate that groundwater levels at the site will likely be consistent with the water level in Rock Creek and could fluctuate during the wet and dry seasons.

4.0 CONCLUSIONS

In our opinion, the proposed bridge is feasible, provided the site is prepared as recommended in this report. We recommend the bridge be supported on driven piles to depths between 56 and 89 feet below the bottom of cap elevation. MSE walls, approach embankments, and detention pond retaining walls will be feasible, provided site preparation is performed as outlined in this report. We recommend that MSE and CIP retaining walls be founded on 2 feet of stone embankment subgrade stabilization. The primary geotechnical considerations for the project are summarized in the “Executive Summary.” Our design and construction recommendations for the project are provided in the following sections.

5.0 DESIGN

5.1 PERMANENT SLOPES

Permanent fill slopes on the site should not exceed a gradient of 2H:1V, unless specifically evaluated for stability. Slopes should be planted with appropriate vegetation or armored to provide protection against erosion as soon as possible after grading. Surface water runoff should be collected and directed away from slopes to prevent water from running down the face of the slope.

5.2 SEISMIC CONSIDERATIONS

5.2.1 Geologic Hazards

5.2.1.1 General

In addition to ground shaking, site-specific geologic conditions can influence the potential for earthquake damage. Deep deposits of loose or soft alluvium can amplify ground motions, resulting in increased seismic loads on structures. Other geologic hazards are related to soil failure and permanent ground deformation. Permanent ground deformation could result from liquefaction, lateral spreading, landsliding, and fault rupture. The following sections provide discussion regarding potential seismic hazards.

5.2.1.2 Fault Surface Rupture

Fault surface rupture is a seismic hazard that includes abrupt, large, differential ground movements that can cause damage to structures located on a known fault. Based on Personius (2017), the closest mapped fault is the Helvetia fault, which is located approximately 3.1 miles north of the site. Due to the distance of the fault, the risk of fault rupture is low at the bridge site.

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5.2.1.3 Liquefaction and Lateral Spreading

Liquefaction can be defined as the sudden loss of shear strength in a soil due to an excessive buildup of pore water pressure. Liquefied soil layers generally follow a path of least resistance to dissipate pore pressures, often resulting in sudden surface settlement, sand boils or ejections, and/or lateral spreading in extreme cases. Clean, loose, uniform or silty, fine-grained, saturated sand is particularly susceptible to liquefaction.

Based on the anticipated depth to groundwater and the subsurface conditions observed below the groundwater level, it is our opinion that the potential for liquefaction at the site is low. As the potential for liquefaction is low, the potential for lateral spreading is low and not considered a hazard for the site. Based on the results of cyclic direct shear testing, the silt and clay at the site will experience a 10 to 15 percent reduction in shear strength due to earthquake loading.

5.2.2 General

According to the ODOT GDM (2018), new bridges west of U.S. Highway 97 are to be designed for a two-level seismic design criterion. The criteria are presented below:

“Life-Safety” Design Criteria:

Under this level of shaking, the bridge and approach structures, foundations, and approach fills must be able to withstand the design forces and displacements without collapse of any portion of the structure and also be consistent with the Life Safety seismic design criteria described below and in the current ODOT GDM. In general, bridges that are properly designed and detailed for seismic loads can accommodate relatively large deflections without the danger of collapse.

Ground motions for this criterial have an average return period of 1,000 years (7 percent probability of exceedance in 75 years). This level of service is typically what has been provided for bridge design prior to the 2018 version of the ODOT GDM.

“Operational” Design Criteria:

This criterion is associated with bridges remaining in service following a level of ground shaking associated with a full-rupture CSZE. For the Operational performance level, bridges and approach fills are designed to provide access for emergency vehicles associated with a full rupture CSZE.

Some structural damage is anticipated, but the damage should be repairable and the bridge should be able to carry emergency vehicles immediately following the earthquake. This holds true for the approach fills leading up to the bridge.

Approach fill settlement and lateral displacements should be minimal to provide for immediate emergency vehicle access for at least one travel lane. As a general rule, an estimated lateral embankment displacement of up to 1 foot is considered acceptable in many cases as long as the “operational” performance criteria described above can be met and the structure foundations are adequately designed to withstand the soil loads resulting from the lateral displacements. Vertical settlements on the order of 6 to 12 inches may be acceptable depending on the roadway geometry, anticipated performance of the bridge end panels, and the ability of bridge foundation elements to withstand any imposed downdrag loads.

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5.2.3 Seismic Design Parameters

Earthquake ground motions at the site were evaluated using seismic data from the 2016 ODOT Seismic Design maps. Peak bedrock accelerations for return periods of 500 years and 1,000 years for the “Life Safety” criteria are provided in Table 1. Based on geologic mapping and our explorations, the recommended seismic design Site Class is D. The “Life Safety” seismic design parameters for Site Class C soil are provided in Table 2.

Table 1. ODOT 2016 Seismic Design Parameters for Bedrock “Life Safety” Criteria

Parameter Value

Site Class (Bedrock) B

Return Period 500 years 1,000 years

PGA 0.180 g 0.283 g

Ss 0.390 g 0.619 g

S1 0.137 g 0.230 g

Table 2. ODOT 2016 Seismic Design Parameters for Soil “Life Safety” Criteria

Site Class (Soil) D

Return Period 500 years 1,000 years

FPGA 1.439 1.318

Fa 1.488 1.311

Fv 2.326 2.140

SDS 0.580 g 0.802 g

SD1 0.319 g 0.493 g

As 0.260 g 0.372 g

Peak bedrock accelerations for the ”Operational” criteria are provided in Table 3. Based on geologic mapping and our explorations, the recommended seismic design Site Class is D. The “Operational” seismic design parameters for Site Class D soil are provided in Table 4.

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Table 3. ODOT 2016 Seismic Design Parameters for Bedrock “Operational” Criteria

Site Class (Bedrock) B

Return Period CSZE

PGA 0.16 g

Ss 0.33 g

S1 0.18 g

Table 4. ODOT 2016 Seismic Design Parameters for Soil “Operational” Criteria

Site Class (Soil) D

Return Period CSZE

FPGA 1.48

Fa 1.536

Fv 2.24

SDS 0.507 g

SD1 0.403 g

As 0.237 g

5.3 BRIDGE FOUNDATION SUPPORT

5.3.1 General

Based on the results of our explorations at the site, the near-surface soil at the bridge generally consists of soft to medium stiff silt and clay catastrophic flood deposits overlying stiff to very stiff silt and clay deposits of the Sandy River Mudstone equivalent. Portions of the near-surface soil are loose or soft; accordingly, conventional spread footings are not an appropriate foundation type and deep foundations will be required to support the proposed bridge.

Based on our experience, the soil conditions encountered in the explorations, and the anticipated loads, driven piles will be the most cost-effective deep foundation for the project.

Driven piles can be installed relatively quickly, produce no spoils, and will generally be the cleanest of the deep foundation alternatives. In addition, testing of the piles can be completed to achieve lower resistance factors than other deep foundation alternatives.

Based on our experience with the County, 16-inch-diameter, closed-end driven pipe piles are the preferred deep foundation type for bridges. We have provided resistances for these piles in the following sections.

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5.3.2 Driven Pile Resistances

Based on the soil encountered in our explorations, 16-inch-diameter, closed-end pipe piles may be used to support the proposed bridge. The maximum service, strength, and extreme event loads for the pile bents are provided in Table 5. The ground surface, top of pile cap, and bottom of pile cap elevations for the bridge bents are provided in Table 6.

Table 5. Maximum Strength, Service, and Extreme Event Pile Loads

Pile Loads Bent Number

Bent 1 Bent 2 Bent 3 Maximum Factored Strength V Compression Load per Pile (kips) 191 195 191

Maximum Extreme Event Compression Load per Pile (kips) 274 471 274

Table 6. Pile Cap Elevations

Bent Number

Ground Surface Elevation

(feet)

Top of Pile Cap Elevation

(feet)

Bottom of Pile Cap Elevation

(feet) Bent 1 139 132.96 127.96 Bent 2 137 130.29 125.29 Bent 3 139 133.96 128.96

We completed an analysis to determine factored axial resistances for individual piles using methods described in AASHTO LRFD (2020). Our analysis is based on the following design and construction assumptions:

The factored pile resistances are based on a resistance factor of 0.65 associated with the use of dynamic pile testing (PDA) and a calibrated wave equation analysis during construction. This corresponds to an approximate factor of safety of 1.54 for soil resistance on the piles. Without dynamic pile testing, AASHTO LRFD (2020) requires that a significantly lower resistance factor of 0.4 be used (an equivalent factor of safety of approximately 2.5), which will result in longer piles.

The axial resistance factors of the piles should meet the requirements of Section 6.5.4.2 of

AASHTO LRFD (2020).

Downdrag effects are included on the abutment piles (Bents 1 and 3) from the maximum approximately 23 feet of fill that will be required for bridge approach construction.

The estimated scour depths at the bent locations are all above the bottom of cap elevations;

therefore, scour effects were not included in our analyses.

The minimum pile spacing, clearance, and embedment into the cap as described in this section are adhered to.

According to AASHTO LRFD (2020), 0.4 inch of settlement is required to develop downdrag on deep foundations. As described in the “Bridge Approach and Detention Pond Fills” section, new approach fills up to 23 feet in height will be required near the proposed bridge abutments and will result in downdrag on the piles. The resistance plots in Appendix C include downdrag effects.

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Table 7 provides the factored downdrag loads that will be exerted at individual foundations at each abutment. The downdrag loads include a load factor of 1.05. The factored downdrag load should be included in the structural capacity of the piles.

Table 7. Settlement-Induced Downdrag Loads

Bent Number Downdrag Depth

(feet) Downdrag Load

(kips) Bent 1 50 86 Bent 2 0 0 Bent 3 55 106

Downdrag loads were applied to the factored loads for Bents 1 and 3. Downdrag loads will not be present at Bent 2. Strength loads should have a resistance factor of 0.65 applied for pile resistances verified by dynamic pile testing, and a resistance factor of 1.0 should be applied to the extreme event loads. Tables 8 and 9 provide estimated minimum tip elevations, estimated minimum tip lengths, and the required axial resistances per pile assuming PDA testing will be conducted on two piles per bent. Pile resistance capacity plots for the bridge bents are provided on Figures C-1 through C-6 in Appendix C. The plots present the factored axial resistances for the maximum strength limit state load for single piles and the nominal axial resistances for the maximum extreme event limit state load for a single pile. Analysis verified that the service limit loading will induce settlements less than described in the “Deep Foundation Settlement” section.

Table 8. Strength Limit State Pile Capacities and Lengths for 16-Inch-Diameter, Closed-Ended Pipe Piles

Total LRFD- Factored

Load1 (kips)

Total Nominal Geotechnical Resistance2

(kips)

Bottom of Pile Cap Elevation

(feet)

Estimated Tip

Elevation (feet)

Estimated Pile Length Below

Pile Cap Elevation

(feet) Bent 1 277 426 127.96 42 86 Bent 2 195 300 125.29 55 70 Bent 3 297 457 128.96 39 90

1. Factored strength load + factored downdrag load

2. Includes a resistance factor of 0.65.

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Table 9. Extreme Event Limit State Pile Capacities and Lengths for 16-Inch-Diameter, Closed-Ended Pipe Piles

Total Nominal Extreme Event Geotechnical

Resistance1 (kips)

Bottom of Pile Cap Elevation

(feet)

Estimated Tip Elevation

(feet)

Estimated Pile Length Below Pile Cap

Elevation (feet)

Bent 1 274 127.96 58 70 Bent 2 471 125.29 39 87 Bent 3 274 128.96 52 77

1. Resistance factor = 1.0

Based on the results of our pile capacity calculations, the strength limit state loads including downdrag will control for Bents 1 and 3, and the extreme event load will control at Bent 2. We have provided pile factored and nominal resistance plots for Bents 1 through 3 of the proposed bridge in Appendix C.

5.3.3 Wave Equation Method

A suitable hammer shall be selected for driving piles and for performance of wave equation analyses. The 1987 or newer version of WEAP should be used. The wave equation analyses should be conducted by personnel qualified by training and experience to perform this type of work. The analysis should be completed using a triangular distribution.

The WEAP analysis should be concurrent and submitted with the Pile and Driving Equipment Data Form. The Agency will approve or reject the pile driving equipment submittal after a review of the wave equation analysis conducted by the contractor. The pile driving hammers should meet the following requirements based on wave equation analysis:

1. The energy of the submitted hammer shall produce a wave equation-predicted blow count between 2 and 10 blows per 1 inch for the ultimate capacities, pile lengths, and other conditions specified.

2. The pile stresses indicated by the wave equation analysis at the ultimate pile bearing capacity shall not be greater than the stress at the point of impending damage to the pile. Tensile and compressive stresses in the pile shall be limited to 90 percent of the pile material’s yield strength for the grade of steel specified at any time during pile installation.

Hammers not meeting these requirements should be rejected. The rejected hammers should be replaced with suitable hammers. Use input values for the wave equation analyses according to Table 10.

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Table 10. Wave Equation Analyses Parameters

Pile Type

Pile Length2

(feet)

Quake (inches)

Damping (seconds/foot)

% Skin

(ITYS)

Rndr3 (kips)

Skin Toe Skin Toe

Bent 1 PP 16X0.5 86 0.1 0.26 0.2 0.15 60 426 Bent 2 PP 16X0.5 87 0.1 0.26 0.2 0.15 67 471 Bent 3 PP 16X0.5 90 0.1 0.26 0.2 0.15 55 457

1. Includes factored downdrag loads.

2. The pile length shown does not include the stick-up length above bottom of pile cap.

3. The nominal driving resistance, Rndr.

5.3.4 Deep Foundation Settlement

We anticipate settlement of foundations meeting the requirements above and embedded in the Sandy River Mudstone equivalent stiff to very stiff silt and clay will be less than 1 inch.

5.3.5 Deep Foundation Group Efficiency

The efficiency of deep foundation groups was evaluated for the pile foundations spaced at 4 feet on-center due to overlapping stresses from adjacent piles. Based on the project plans, the proposed 16-inch-diameter pipe pile foundation spacing will be 3B (diameter), center to center, negating group effects to the allowable axial capacity. We anticipate the pile caps will be constructed in firm contact with the ground.

We have evaluated the group effect of lateral loads for each bridge bent using the program GROUP – 2019 version 11.10 by Ensoft, Inc. The soil parameters for our group lateral analysis are provided in Tables 11 through 13.

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Table 11. GROUP Input Parameters for Bent 1

Soil

GROUP

Soil Type

Effective Unit

Weight, ' (pcf)

Undrained Cohesion

(psf)

Friction Angle, Strain Factor, ε50

Soil Modulus, k (pci)

Medium Stiff Silt (0 to 7 feet BGS)

Soft Clay 110 250 - 0.02 -

Very Stiff Silt (7 to 16 feet BGS)

Stiff Clay with Free

Water 48 1,500 - 0.01 1,000

Medium Stiff to Stiff Silt

(16 to 55 feet BGS)

Stiff Clay with Free

Water 48 750 - 0.004 500

Stiff Silt (55 to 80 feet BGS)

Stiff Clay with Free

Water 48 1,000 - 0.004 700

Dense Sand (80 to 140 feet)

Sand (Reese)

53 - 36 - 60

Table 12. GROUP Input Parameters for Bent 2

Soil Type

Effective Unit

Weight, ' (pcf)

Undrained Cohesion

(psf)

Friction Angle, Strain Factor, ε50

Soil Modulus, k (pci)

Soft to Medium Stiff Silt

(0 to 5 feet BGS) Soft Clay 110 250 - 0.02 -

Very Stiff Silt (5 to 15 feet BGS)

Stiff Clay with Free

Water 48 2,000 - 0.004 400

Medium Stiff to Stiff Silt

(15 to 55 feet BGS)

Stiff Clay with Free

Water 48 1,000 - 0.005 200

Stiff Silt (55 to 80 feet BGS)

Stiff Clay with Free

Water 48 2,000 - 0.004 400

Dense Sand (80 to 140 feet)

Sand

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Table 13. GROUP Input Parameters for Bent 3

Soil Type

Effective Unit

Weight, ' (pcf)

Undrained Cohesion

(psf)

Friction Angle, Strain Factor, ε50

Soil Modulus, k (pci)

Soft to Medium Stiff Silt

(0 to 5 feet BGS) Soft Clay 110 250 - 0.02 -

Very Stiff Silt (5 to 15 feet BGS)

Stiff Clay with Free

Water 48 1,000 - 0.004 1,000

Medium Stiff to Stiff Silt

(15 to 35 feet BGS)

Stiff Clay with Free

Water 48 750 - 0.005 500

Stiff Silt (35 to 70 feet BGS)

Stiff Clay with Free

Water 48 1,000 - 0.004 750

Medium Stiff to Stiff Silt

(70 to 120 feet BGS)

Stiff Clay with Free

Water 48 750 - 0.005 500

Dense Sand (120 to 140 feet)

Sand

The bent loads and pile cap layout were provided by NV5’s structural group. The bent loads are summarized in Table D-1 in Appendix D. The output files for our GROUP analyses are presented in Appendix D.

5.4 SIGNAL POLE FOUNDATION DESIGN

We understand that mast-arm signal pole will be installed for a new signal at the intersection of SE Lois Street and Century Boulevard. The signal pole will be constructed in accordance with ODOT standards. The design of traffic control structures is subject to the requirements of Chapter 16 to the ODOT GDM (2018) and Section 13 of AASHTO (2015).

We anticipate the foundation for the signal pole will consist of 36- to 48-inch-diameter drilled shafts. We anticipate that the axial loads on the structure will be negligible and the overturning and torsional forces will control design.

We understand that LPILE p-y curves will be used to evaluate the pole foundation; deflection of the foundation should be limited to ½ inch at the shaft head. The appropriate soil parameters for the LPILE design method are provided in Table 14.

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Table 14. LPILE Input Parameters for Signal Pole Design

LPILE

Soil Type

Effective Unit

Weight, ' (pcf)

Undrained Cohesion

(psf)

Friction Angle, Strain Factor, ε50

Soil Modulus, k (pci)

Medium Stiff Silt (0 to 7.5 feet BGS)

Stiff Clay w/o Free

Water 110 750 - 0.01 100

Medium Stiff Silt (7.5 to 16 feet BGS)

Stiff Clay with Free

Water 48 750 - 0.01 100

Loose Sand (16 to 21 feet)

Sand (Reese)

53 - 30 - 20

Medium Stiff Silt (21 to 27 feet)

Stiff Clay with Free

Water 48 750 - 0.01 100

Very Stiff Silt (27 to 31.5 feet)

Stiff Clay with Free

Water 48 2,000 - 0.005 1000

5.5 BRIDGE APPROACH AND DETENTION POND FILLS

5.5.1 General

MSE walls are proposed for retaining the bridge approach embankments. We understand that the MSE wall design will be the responsibility of the contractor and will be selected from a preferable proprietary MSE wall system as described on the project plans. GeoDesign/NV5 is responsible for completing the global stability analysis for the wall design. The results of our overall stability or global stability analysis of design details are presented in the “Retaining Structures” section. Global stability analysis should be verified when final design of the MSE walls is completed.

5.5.2 Fill-Induced Settlement

The proposed approach fills for MSE walls and embankments will cause consolidation of the soil underlying the planned walls and embankments. We completed analyses to determine the potential settlement associated with the proposed fills for the project. Our settlement estimates are based on subsurface information obtained from our borings, the results of our laboratory testing, and our experience in the site area.

Based on information provided by NV5, the approach MSE fill walls will have maximum heights of 21 feet for Wall 1 and 23 feet for Wall 2. A detention pond will be constructed on the east side of the south approach. A CIP retaining wall will be constructed on the north side of the detention pond to retain fills adjacent to the wetlands of Rock Creek. The detention area will have a maximum fill height of 12 feet above existing grades.

Based on our analysis, fill-induced settlement of the south approach (Wall 1) will be between 3 and 4 inches. The detention pond fills will induce settlements ranging from 1 inch to 2 inches.

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The north approach (Wall 2) will induce settlements between 4 and 5 inches. The fill-induced settlement of the CIP retaining wall will range from 1 inch to 2 inches. Fill placed for the detention pond will not add to the settlement of the south bridge approach because the fill height in the detention area is less than the Wall 1 height and will not increase the effective stress below the bridge approach.

5.5.3 Primary Consolidation Time Considerations

The soil conditions below the embankment fill areas generally consist of fine-grained material.

The fine-grained silt and clay can result in relatively long consolidation times. Based on our analysis, we anticipate that approximately three-quarters of the embankment settlement will occur within eight to ten weeks; however, because of the fine-grained soil, the time required to complete the last one-quarter of the settlement will be approximately four months. The construction of pavement sections and any other structures in fill and wall areas should be delayed until primary consolidation is complete.

We recommend that survey monuments be placed on the bridge approach MSE walls after wall construction is complete. The survey monuments should be installed and immediately surveyed following completion of wall construction. The monuments should be surveyed at least twice per month using survey equipment with an accuracy of 1/100th of a foot and referenced to a stationary datum established at least 500 feet from fill placement. Once the survey data indicates that the primary settlement has essentially completed, settlement-sensitive structural elements can be constructed on the MSE wall approach fills.

5.6 RETAINING STRUCTURES

5.6.1 General

Based on information provided by NV5, we understand that MSE walls are planned to support the fills for the bridge approach embankments. Wall 1 on the south side of the bridge will have a maximum height of 21 feet above final grade with a total length of 580 feet. Wall 2 on the north side of the bridge will have a maximum height of 23 feet above final grade with a total length of 386 feet. A CIP concrete retaining wall will be constructed on the east side of the south approach MSE wall (Wall 1) to retain fill slopes for a stormwater detention pond with a maximum retained height of 12.7 feet. Weak and compressible silt and clay were encountered in the upper 5 to 8 feet of the soil profile across the site. Therefore, we recommend that the MSE and CIP walls be founded on a minimum thickness of 2 feet of stone embankment sub-excavation material.

5.6.2 Retaining Wall Design Recommendations

5.6.2.1 MSE Wall Design Parameters

Our MSE retaining wall design recommendations are based on the following assumptions: (1) the walls consists of MSE walls, (2) the walls will be less than 23 feet in height, and (3) the appropriate wall surcharges are included in the design as described in this section.

MSE walls should be designed using the soil parameters provided in Table 15. The material used to backfill behind all MSE walls is discussed in the “Materials” sections.

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Table 15. MSE Wall Design Parameters

Soil Type Unit

Weight (pcf)

Friction Angle, ϕ (degrees)

Cohesion (psf)

Retained Soil – Reinforcement Zone (Granular Wall Backfill) 1

125 36 0

Foundation Soil (2 Feet of Stone Embankment Subgrade

Stabilization Overlying Native Soil)2 110 28 0

1. Wall backfill should meet the specifications provided in OSSC 00510.12 – Granular Wall Backfill.

2. Foundation soil should consist of 2 feet of stone embankment stabilization material prepared in conformance with the “Site Preparation” section.

5.6.2.2 CIP Wall Design Parameters

Our CIP retaining wall design recommendations are based on the following assumptions: (1) the walls consists of CIP walls, (2) the walls will be less than 12.7 feet in height, and (3) the appropriate wall surcharges are included in the design as described in this section.

CIP walls should be designed using the soil parameters provided in Table 16. The material used to backfill behind all walls is discussed in the “CIP Wall Backfill and Drains” and “Materials” sections.

Table 16. CIP Wall Design Parameters

Soil Type Unit

Weight (pcf)

Friction Angle, ϕ (degrees)

Cohesion (psf)

Retained Soil 125 34 0 Foundation Soil

(Granular Fill Overlying Native Silt and Clay)1 125 30 0

1. Wall backfill should meet the specifications provided in OSSC 00510.12 (Granular Wall Backfill).

2. Wall foundation subgrade should consist of 2 feet of granular backfill subgrade stabilization prepared in conformance with the “Site Preparation” section.

5.6.2.3 Wall Surcharges

Where traffic loads are located within a horizontal distance from the top of the wall equal to one-half the wall height (not anticipated), the lateral earth pressure shall be increased by a surcharge load equal to 2 feet of soil (assuming a soil density of 125 pcf).

The design equivalent fluid pressure should be increased for walls that retain sloping soil. We recommend the following lateral earth pressures be increased using the following factors (Table 17) when designing walls that retain sloping soil.

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Table 17. Lateral Earth Pressure Increase Factors for Slope Soil Backfill

Slope of Retained Soil

(degrees) Lateral Earth Pressure

Increase Factor

0 1.00 5 1.06

10 1.12 20 1.33 25 1.52 30 2.27

Seismic forces should be modeled based on the pseudo-static approach developed by the Mononobe-Okobe method. We recommend using a seismic coefficient of one-half of the PGA when analyzing internal stability. The PGA should be modified to account for potential soil amplification using the seismic site coefficients described in “Seismic Considerations” section.

5.6.2.4 Wall Global Stability

Wall global stability analyses were performed for the proposed retaining walls at the northern and southern bridge approaches assuming MSE-type retaining walls will be constructed for Walls 1 and 2 and a CIP retaining wall will be constructed to retain detention pond fills east of the south bridge approach. We understand that MSE retaining walls will be constructed around the perimeter of the bents. The global stability analysis was performed using a limit-equilibrium analysis using SLOPE/W software at the most critical section at each wall location. Our analyses of the selected wall types were performed based on roadway alignment, existing ground surface topography, and embedment depths shown on the drawings provided by NV5. For design purposes, we have assumed that the MSE retaining walls will be constructed with minimum reinforcement lengths equal to 0.7*H (0.7 times the height of the wall) or 8 feet, whichever is greater, and a geogrid reinforcement with a minimum allowable design strength of 1,450 pounds per foot.

A live load of 250 psf was used for traffic surcharge loading. The retaining walls were analyzed under static and seismic conditions. For pseudo-static seismic conditions, we used a horizontal coefficient (kh) of 0.19. We have also performed a post-seismic analysis using a lower Kh value of 0.008, which will occur near the end of earthquake loading after foundation soil has experienced strain softening of the soil. Based on our cyclic direct shear testing, the silt and clay at the site will experience a 10 to 15 percent reduction in shear strength due to earthquake loading. Our retaining wall global stability results are provided in Table 18.

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Table 18. Summary of Retaining Wall Global Stability Results

Wall Station Geogrid Length (feet)

Global Stability FOS Static Conditions

Global Stability FOS Pseudo-Static

Conditions

Global Stability FOS Post-Seismic

Conditions

Wall 1 Station 2+40

20 1.71 1.21 1.10

Wall 2 Station 3+00

20 1.72 1.24 1.11

CIP Wall Station 1+04

- 2.13 1.42 1.84

The results of our slope stability analyses are presented in Appendix E.

5.6.2.5 MSE Wall Design Parameters

The following retaining wall design parameters have been established for MSE Walls 1 and 2.

00596A – MSE Walls Foundation soil unit density ............................................................... 0.110 kips per square foot Foundation soil angle of internal friction ..................................................................... 28 degrees Foundation soil nominal (unfactored) bearing resistance .................. 12.5 kips per square foot Retained soil unit density ................................................................................................... 125 pcf Retained soil angle of internal friction......................................................................... 36 degrees Reinforced soil unit density ................................................................ 0.125 kips per square foot Reinforced soil angle of internal friction ..................................................................... 36 degrees Peak ground acceleration coefficient (PGA) ........................................................................ 0.28 g Short period spectral acceleration coefficient (SS). ............................................................ 0.62 g Long period spectral acceleration coefficient (S1) .............................................................. 0.23 g Site class ........................................................................................................................................ D Peak seismic ground acceleration coefficient modified by zero period site factor (As)....... 0.37 Horizontal seismic acceleration coefficient (Kh) .................................................................... 0.19

5.6.2.6 Temporary Cuts

Temporary cuts may be required to construct the proposed retaining walls. Excavations into the slopes need to be carefully planned so as not to destabilize the slopes. Cuts less than 4 feet should stand vertical. Deeper excavations should initially be cut back at an inclination 1.5H:1V or flatter or be shored. Depending on the depth of cuts and the soil encountered, excavation inclinations may be able to be steepened to 1.25H:1V or even 1H:1V.

GeoDesign/NV5 should be contacted at the time of excavation to observe the soil and determine if the steeper slope excavations are feasible. Because of the adverse effects that water can have on slope stability, we anticipate that this process will not be feasible for wintertime construction and wall construction should occur in dry summer months or temporary shoring used.

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5.6.2.7 CIP Wall Foundations

Wall foundations may be supported on firm native soil or organic-free fill soil, provided they are evaluated as described in the “Site Preparation” section. Footings supporting the proposed retaining walls should be designed based on the nominal bearing pressures provided in Table 19.

Table 19. CIP Wall Nominal Bearing Resistance

Wall Station Nominal Bearing Resistance

(ksf)

Station 0+00 to 0+25 7,950 Station 0+25 to 1+40 7,025 Station 1+40 to 1+60 7,900

5.6.2.8 Resistance to Sliding

Lateral loads on footings can be resisted by passive earth pressure on the sides of the structures and by friction on the base of the footings. Our analysis indicates that the available passive earth pressure for footings is 250 pcf. Adjacent footing elements, pavement, or the upper 12-inch depth of adjacent, unpaved areas should not be considered when calculating passive resistance.

A coefficient of friction equal to 0.60 may be used when calculating resistance to sliding for…