Geotech_Appdx_LSAR_Reach9_Phase_5A_Oct2014.pdf

Los Angeles District

Geotechnical Branch

Geotechnical Appendix to Design Documentation Report, Geotechnical Investigation for the Lower Santa Ana River

Reach 9 Phase 5A Yorba Linda, California

Prepared By:

October 09, 2014

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TABLE OF CONTENTS

SECTION PAGE

1.0 INTRODUCTION .................................................................................... 1-1

1.1. PROJECT DESCRIPTION ................................................................................................................... 1-1

1.2. PURPOSE AND SCOPE OF INVESTIGATION ........................................................................................ 1-2

2.0 GEOTECHNICAL INVESTIGATION ...................................................... 2-1

2.1 INITIAL GEOTECHNICAL INVESTIGATION ........................................................................................... 2-1

2.2 SUPPLEMENTAL GEOTECHNICAL EXPLORATIONS ............................................................................. 2-1

2.2.1 Cone Penetration Testing ........................................................................................................ 2-2

2.2.2 Test Holes ................................................................................................................................ 2-3

2.2.3 Laboratory Testing ................................................................................................................... 2-4

2.2.4 Slug Testing ............................................................................................................................. 2-4

3.0 SITE AND GEOLOGIC CONDITIONS ................................................... 3-1

3.1 GEOLOGIC AND TECTONIC SETTING ................................................................................................. 3-1

3.2 SURFACE CONDITIONS .................................................................................................................... 3-2

3.3 SUBSURFACE CONDITIONS .............................................................................................................. 3-2

3.4 GROUNDWATER .............................................................................................................................. 3-3

4.0 SEISMIC AND GEOLOGIC HAZARDS .................................................. 4-1

4.1 FAULT RUPTURE ............................................................................................................................. 4-1

4.2 STRONG GROUND SHAKING ............................................................................................................ 4-1

4.3 SOIL LIQUEFACTION ........................................................................................................................ 4-3

4.4 STABILITY OF SLOPES PROTECTED BY GROUTED RIP RAP ............................................................... 4-6

4.5 SEISMIC DEFORMATION ANALYSES ................................................................................................. 4-8

5.0 GEOTECHNICAL DESIGN RECOMMENDATIONS .............................. 5-1

5.1 EARTHWORK ................................................................................................................................... 5-1

5.1.1 Site Preparation ....................................................................................................................... 5-1

5.1.2 Fill and Backfill ......................................................................................................................... 5-1

5.2 SHEET PILE WALLS ......................................................................................................................... 5-2

5.2.1 Geotechnical Parameters ........................................................................................................ 5-2

5.2.2 Tieback Anchors ...................................................................................................................... 5-6

5.3 SIDE DRAIN EXTENSIONS ................................................................................................................ 5-6

5.4 CORROSION .................................................................................................................................... 5-7

6.0 CONSTRUCTION CONSIDERATIONS .................................................. 6-1

6.1 EARTHWORK ................................................................................................................................... 6-1

6.1.1 Excavation Characteristics ...................................................................................................... 6-1

6.1.2 Excavations ............................................................................................................................. 6-1

6.1.3 Dewatering ............................................................................................................................... 6-1

6.2 SHEET PILE INSTALLATION .............................................................................................................. 6-2

6.3 TIEBACKS ANCHORS ....................................................................................................................... 6-2

6.4 INSTRUMENTATION AND MONITORING .............................................................................................. 6-2

6.4.1 Survey Monuments .................................................................................................................. 6-3

6.4.2 Inclinometers ........................................................................................................................... 6-3

7.0 LIMITATIONS ......................................................................................... 7-1

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8.0 REFERENCES ....................................................................................... 8-1

List of Tables

Table 1 – CPT Advancement Depths Table 2 – Summary of Test Hole Depths and Elevations Table 3 – Fault Model Parameters Table 4 – Peak Ground Accelerations for OBE and MDE Table 5 – Spectral Accelerations for 5% Damping Table 6 – Zones of Potentially Liquefiable Soils Table 7 – Recommended Minimum Factors of Safety for Slope Stability Analyses Table 8 – Summary of Slope Stability Results Table 9 – Summary of Laboratory Direct Shear Results Table 10 – Recommended Soil Parameters for the Design of Sheet Pile Walls (using CWALSHT and CWALSSI Programs) Table 11 – Recommended Geotechnical Parameters for the Design of Side Drain Extensions

List of Figures

Figure 1 - Vicinity Map Figure 2 – Site Plan Figure 3 – Geologic Map Figure 4 – Regional Fault and Epicenter Map Figure 5 – Geologic Sections A-A’ through E-E’ Figure 6 – Geologic Section F-F’ Figure 7 – Probabilistic Response Spectra Figure 8 – Slope Stability –Sta 1235+10; Temporary Backcut Figure 9 – Slope Stability –Sta 1239+60; Static, Full Scour Case Figure 10 – Slope Stability –Sta 1235+10; Pseudo-Static (OBE), Full Scour Case Figure 11 – Slope Stability –Sta 1239+60; Pseudo-Static (MDE), Full Scour Case Figure 12 – Slope Stability –Sta 1235+10; Pseudo-Static (MDE), No Scour Case Figure 13 – Slope Stability –Sta 1239+60; Post Liquefaction Case Figure 14 – Slope Stability Sta 1235+10; Rapid Drawdown, Full Scour Case Figure 15 – Slope Stability –Sta 1239+60; Temporary Backcut Figure 16 – Slope Stability –Sta 1235+10; Static, Full Scour Case Figure 17 – Slope Stability –Sta 1239+60; Pseudo-Static (OBE), Full Scour Case Figure 18 – Slope Stability –Sta 1235+10; Pseudo-Static (MDE), Full Scour Case Figure 19 – Slope Stability –Sta 1239+60; Pseudo-Static (MDE), No Scour Case Figure 20 – Slope Stability –Sta 1235+10; Post Liquefaction Case Figure 21 – Slope Stability –Sta 1239+60; Rapid Drawdown, Full Scour Case

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Attachments

Attachment A Explorations and Laboratory Test Results (USACE 2013) Attachment B Supplemental Geotechnical Explorations (URS, 2014) Attachment C Laboratory Testing Attachment D Slug Testing Attachment E Calculations

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Acronyms and Abbreviations

ASTM American Society for Testing and Materials, now known as ASTM International bgs below ground surface CDMG California Division of Mines and Geology CGS California Geologic Survey CPT Cone Penetration Test CRR Cyclic Resistance Ratio CSR Cyclic Stress Ratio DDR Design Documentation Report g acceleration due to gravity kips/in kilopounds per inch km kilometer ksf kilopounds per square foot LSAR R9P5A Lower Santa Ana River, Reach 9 Phase 5A mm/yr millimeter/year MDE Maximum Design Earthquake MSE Mechanically Stabilized Earth NGA Next Generation Attenuation NGVD29 National Geodetic Vertical Datum of 1929 OBE Operating Basis Earthquake OSHA Occupation and Safety Hazard Administration PGA Peak Ground Acceleration PSHA Probabilistic Seismic Hazard Analysis RCP Reinforced Concrete Pipe SCEC Southern California Earthquake Center SPT Standard Penetration test TH Test Hole URS URS Group, Inc.

USA Underground Service Alert USACE U.S. Army Corps of Engineers USGS United States Geologic Survey

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1.0 INTRODUCTION

This report presents the results of URS Group, Inc.’s (URS) geotechnical investigation for the proposed bank protection along the Lower Santa Ana River, Reach 9 Phase 5A (LSAR R9P5A), in Yorba Linda, California. This geotechnical investigation supplements an initial geotechnical investigation completed by the U.S. Army Corps of Engineers (USACE) in 2013. The results of engineering analyses of the proposed bank protection, and geotechnical recommendations for final design, are also included in this report.

The project area is located along the north bank of the Santa Ana River parallel to East La Palma Avenue and just upstream of Yorba Linda Boulevard. The LSAR R9P5A alignment is approximately 4,150 feet in length. The area is geographically known as Horseshoe Bend as shown on USGS 7.5 minute quadrangle topographic maps. The location of the project area is shown on the Vicinity Map, Figure 1.

1.1. PROJECT DESCRIPTION

The riverbank along the northern edge of the Santa Ana River in the project area supports a paved two-way bike and pedestrian path. The riverbank extends about 15 feet or more above the river channel. Existing rip rap armor extends to an unknown toe down depth. The rip rap currently provides scour protection for the bike/pedestrian path and East La Palma Avenue which runs parallel to the riverbank. The current rip rap protection is not sufficient for deeper scour levels that are now anticipated along this reach of the Lower Santa Ana River.

The Phase 5A alignment begins at Station 1233+60 (downstream end) and ends at station 1274+53 (upstream end). The new bank protection that is proposed along the project alignment includes grouted rip rap slope protection beginning at Station 1233+60 and extending upstream along the alignment to Station 1244+60, and anchored sheet piles from Station 1244+57 upstream to Station 1274+53. The sheet piles will be used in the area where the active river channel is very close to the north riverbank. The extents of the grouted rip rap and sheet pile alignments are shown in Figure 2 Site Plan.

The rip rap slope protection will be 24 inches thick (measured horizontally) and extend from the top of the current riverbank down to the anticipated scour depths (about 16 feet below the current river thalweg), at an inclination of 2:1 Horizontal:Vertical (H:V). Temporary excavations will be required to install the rip rap to the depth of the anticipated scour. This will generally result in

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1-2 a total slope protection height of 35 to 40 feet. The rip rap slope protection will also be adjacent to an existing retail center at about Station 1238+00. There is an existing Mechanically Stabilized Earth (MSE) retaining wall with a wall height up to 15 feet along the property line of the retail center, adjacent to where the rip rap is proposed.

The anchored sheet piles will be driven along the top edge of the existing riverbank, parallel to the bike/pedestrian path. The anchor will be located about 8 to 9 feet from the top of the sheet pile. After the maximum scour, it is anticipated that the maximum exposed sheet pile wall height would be on the order of 36 feet.

Side drains are proposed to enter the river at six locations along the LSAR R9P5A alignment.

The Reinforced Concrete Pipe (RCP) side drains range in size from 27 to 60 inches in diameter.

The project will extend these side drains through the proposed rip rap protection or sheet pile.

1.2. PURPOSE AND SCOPE OF INVESTIGATION

The purpose of the geotechnical investigations was to evaluate the subsurface conditions and to provide geotechnical recommendations for design and construction of the proposed LSAR R9P5A riverbank protection. The scope of the investigations included review of available site information; performing subsurface explorations including Cone Penetration Tests (CPTs), test trenches, and test holes (geotechnical borings); installing open standpipe piezometers; slug testing; geotechnical laboratory testing; developing site-specific design ground motions;

performing geotechnical evaluations and analyses; and preparing this geotechnical appendix.

The results of the geotechnical investigations were used to develop the geotechnical discussions, conclusions, and recommendations presented in this appendix regarding the following:

• Site and subsurface conditions;

• Geologic and tectonic setting;

• Potential geologic and seismic hazards, including fault rupture, strong ground shaking, slope stability and soil liquefaction;

• Geotechnical design recommendations for the design of the sheet pile wall, grouted rip rap slope protection and side drain extensions; and

• Construction considerations.

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Detailed results of the previous subsurface investigations, current field explorations, and laboratory testing are provided in the attachments of this appendix.

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2.0 GEOTECHNICAL INVESTIGATION

The geotechnical investigation included reviews of previous geotechnical information and available literature, subsurface explorations including CPT soundings, test trenches and holes, and laboratory testing. The locations of the explorations are shown in Figure 2.

2.1 INITIAL GEOTECHNICAL INVESTIGATION

USACE performed an initial geotechnical investigation for the project in 2013. The investigation included 10 test holes (designated TH13-R9P5A-01 through TH13-R9P5A-10), 13 test trenches (designated TT13-R9P5A-01 through TT13-R9P5A-13), and six CPT soundings (designated CPT-01 through CPT-06). Two of the test holes were converted to open standpipe piezometers (P13-R9P5A-08 and P13-R9P5A-09). The geotechnical testing program included grain-size distribution, plasticity index, moisture content, dry unit weight, shear strength (direct shear), corrosion potential, and R-Value tests. The results of the USACE investigation are presented in Attachment A of this Geotechnical Appendix.

2.2 SUPPLEMENTAL GEOTECHNICAL EXPLORATIONS

The supplemental geotechnical explorations consisted of advancing six CPTs (designated CPT- 001 through CPT-006), drilling ten test holes (designated TH14-R9P5A-01 through TH14- R9P5A-10), drilling and installing an open standpipe piezometer hole (designated P14-R9P5A- 05), and performing slug tests in the new piezometer along with the previous ones constructed by the USACE.

Prior to mobilizing to the site for field activities, the required boring, access, right-of-entry and encroachment permits were obtained from the County of Orange. The exploration locations were cleared of underground utilities by calling Underground Service Alert (USA) 48 hours prior to fieldwork.

All of the exploration locations were located at the toe of the existing riverbank, with the exception of TH14-R9P5A-07 through TH14-R9P5A-10, which were located along the bike trail. Traffic control was managed at those locations using two flagmen, warning signs, and orange traffic delimiters to provide early warning for bicycles and pedestrians to avoid and maintain a safe distance from the operating equipment.

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Environmental noise control was also required for all of the explorations to prevent environmental disturbance adjacent to the project site. Portable acoustic barrier blanket panels were erected adjacent to the exploration equipment to contain equipment noise.

2.2.1 Cone Penetration Testing

On March 6th and 7th, 2014, CPTs were advanced to refusal along the toe of the existing bank protection. The CPTs were primarily advanced downstream of the Via Lomas De Yorba side drain, as only a limited area of the toe upstream of that location was accessible. Figure 2 provides the locations of all the CPTs performed.

A track-mounted CPT rig, provided and operated by Kehoe Testing and Engineering, was used to advance the CPTs. Due to access constraints and the limited downward resistance provided by the anchored CPT rig, it was difficult to advance the CPTs through the dense upper materials encountered along the alignment. The CPTs encountered refusal at relatively shallow depths and multiple attempts were made at several locations. Table 1 outlines the CPTs attempted and their total depths. Detailed logs of the CPTs and field procedures are provided in Attachment B of this Geotechnical Appendix.

Table 1 – CPT Advancement Depths

CPT Total Depth (feet)

CPT-001 22

CPT-002 10

CPT-002B 11

CPT-003 11

CPT-004 5

CPT-004B 5

CPT-005 6

CPT-005B 7

CPT-006 11

CPT-006B 4

CPT-010a 5

Note:

a. The CPTs were generally located adjacent to the test holes. CPT-007 through CPT-009 were not performed;

CPT-010 was performed adjacent to TH14-R9P5A-10.

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2.2.2 Test Holes

The ten test holes (borings) were advanced between March 10 and 24, 2014 to depths ranging from 70 feet to 91 feet below the ground surface (bgs). A URS geologist logged the test holes and obtained alternating disturbed and relatively intact ring samples at about five-foot depth intervals. An additional test hole was drilled to 50 feet bgs adjacent to TH15-R9P5A-05. This test hole was converted to an open standpipe piezometer (P14-R9P5A-05) to monitor seasonal groundwater fluctuation and to perform slug testing as discussed in Section 2.2.4.

The test holes were located between the locations of the previous test holes (USACE, 2013) and were spaced along the proposed sheet pile wall alignment (Figure 2). The test holes were initially advanced using mud rotary drilling. However, shallow ground water and the presence of coarse granular soils caused the holes to cave between sampling intervals. Subsequently, hollow stem augering was employed with drilling mud added to the augers to create a positive head pressure to prevent heaving at the bottom of the augers.

Logs of the borings and piezometer from the supplemental investigation are presented in Attachment B of this report. Table 2 summarizes the depth to the bottom of the test hole, existing ground surface elevation at the test hole location, measured groundwater levels (at the time of drilling), and bedrock elevation for each test hole.

Table 2 – Summary of Test Hole Depths and Elevations

Test Hole No.

Depth to

Bottom of Test Hole (feet, bgs)

Ground Surface

Elevation (feet, NGVD29)

Approximate Elevation of Encountered Ground Water (feet, NGVD29)

Approximate Elevation of

Bedrock (feet, NGVD29)

TH14-R9P5A-01 75.5 336.7 330 264

TH14-R9P5A-02 72.5 336.7 330 269

TH14-R9P5A-03 91.0 332.7 326 262

TH14-R9P5A-04 75.1 335.4 330 270

TH14-R9P5A-05 70.1 337.6 327 272

TH14-R9P5A-06 76.0 337.8 327 275

TH14-R9P5A-07 80.2 351.8 325 274

TH14-R9P5A-08 70.2 351.6 325 284

TH14-R9P5A-09 74.5 352.8 326 284

TH14-R9P5A-10 81.0 355.6 329 278

P14-R9P5A-05 50.0 336.4 325 NE

Note:

NE – Not Encountered

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2.2.3 Laboratory Testing

Geotechnical laboratory testing was completed on representative soil samples to confirm the field classifications and to evaluate the physical and engineering characteristics of the subsurface materials, in accordance with the relevant ASTM International (ASTM) standards.

Representative soil samples were tested for moisture content, dry unit weight, plasticity index, grain size distribution, shear strength (direct shear), and corrosion potential. The test results are shown at the corresponding sample locations on the log of test borings. Graphical results of the laboratory test data are presented as Attachment C.

2.2.4 Slug Testing

URS performed slug tests at piezometers P13-R9P5A-08, P13-R9P5A-09 and P14-R9P5A-05 on April 7, 2014 using a Solinst® levelogger pressure transducer and one barometer (barologger).

The locations of the piezometers are shown on Figure 2. The slug tests were performed to estimate the hydraulic conductivity of the soil deposits in the river channel adjacent to the piezometers. P13-R9P5A-09 was not completely developed at the time of the slug test and the data does not reflect the actual groundwater conditions due to the existence of drilling mud in the piezometer. P13-R9P5A-08 and P14-R9P5A-05 were properly developed (or drilled without drilling mud) and the slug test data collected from these piezometers should appropriately reflect the groundwater conditions in the area of the piezometers.

Static groundwater elevations collected on April 7, 2014 were 330.90, 321.00, and 325.16 feet NGVD29 at P13-R9P5A-08, P13-R9P5A-09, P14-R9P5A-05, respectively. Upon completion of the slug testing, Solinst® levelogger pressure transducers were deployed into each of the three piezometers and one barometer (barologger) was deployed into P14-R9P5A-01. The deployed instruments are setup for long term groundwater measurements. Engineers from the USACE were trained onsite on the use of the monitoring equipment and the long term groundwater level data collection responsibility has been turned over to the USACE.

The initial groundwater level and slug test data are provided in Attachment D.

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3.0 SITE AND GEOLOGIC CONDITIONS

Knowledge of the site and geologic conditions was developed from a review of available information, site reconnaissance, and the findings from the previous and current geotechnical investigations.

3.1 GEOLOGIC AND TECTONIC SETTING

The project site is located within the Santa Ana River floodplain in an area known as Horseshoe Bend. The site is bounded by the Chino Hills to the northwest and by the Santa Ana Mountains to the south. Prado Dam is located approximately 5 miles to the east and upstream of the project area.

The Santa Ana River is incised into a variety of different bedrock materials in the site vicinity and has subsequently been backfilled by Holocene and late Pleistocene age alluvial deposits.

Within the immediate site area, the bedrock is characterized as undifferentiated Puente Formation; sedimentary deposits that may range in age from early Pliocene to Miocene. The Puente formation is sub-divided into four members: Sycamore Canyon Member, Yorba Member, Soquel Member, and the La Vida Member. A geologic map of the site region with local faults is shown in Figure 3.

The project site is in an area characterized by active faults and seismicity. Figure 4 provides a map of active faults in the region. The Elsinore fault zone is a major dextral shear system parallel to the southern portion of the northwest-trending San Andreas fault zone that accommodates about 5 millimeters of slip per year along the Pacific-North American Plate boundary. The Whittier section of the Elsinore fault zone (Whittier fault) is a northwest trending fault system mapped as an Alquist Priolo Earthquake Fault Zone (AP Zone, see Section 4.1) by the State of California. It is a significant seismogenic source in the immediate vicinity of the project site and is located approximately 2,500 feet to the north at its nearest trace.

The Horseshoe fault is a southern branch of a fault that that has been offset by the Whittier fault.

It projects southwestward towards the project site. The fault is referenced in the California Division of Mines and Geology Fault Evaluation Report 41 (CDMG, 1979), and is used as evidence of the Elsinore fault having a right-lateral sense of movement. It is inferred to possibly project under the alluvium near the project area in the older Puente Formation. There is no evidence of movement in the fault within the younger alluvium and it is subsequently not within the AP zone.

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3.2 SURFACE CONDITIONS

The paved Santa Ana River bike trail runs mostly parallel to the bend in the Santa Ana River channel. The surface of the bike trail along the LSAR R9P5A alignment ranges from about elevation 345 feet NGVD29 at the downstream end to about elevation 357 feet NGVD29 at the upstream end. The Santa Ana River flood plain is typically 10 to 15 below the bike trail. The riverbank slope is armored with rip rap with a variable toe down depth. A paved access road down to the flood plain exists at the downstream end of the project (near Station 1238+00) and the upstream end (about 3,500 feet southeast of boring TH14-R9P5A-10). Vehicular access to the toe of the riverbank is currently limited to an area downstream of about Station 1262+50 and an area upstream of Station 1237+00 due to eroded banks. The Santa Ana River channel is primarily densely vegetated with trees and shrubs.

3.3 SUBSURFACE CONDITIONS

The subsurface explorations encountered alluvial deposits overlying bedrock. Fill soils were encountered in the borings performed at the surface of the Santa Ana Bike trail. Figures 5 and 6 provide detailed geologic cross sections along the LSAR R9P5A alignment. The locations of the sections are shown on Figure 2.

Fill materials encountered in the embankment supporting the bike trail were observed to consist of very stiff sandy clay and dense to very dense silty to clayey sand. The fill ranged in depth from 10 feet to 15 feet below the surface of the bike trail.

The alluvial soils encountered consist of interbedded sands, silts and clays. Varying amounts of cobbles were encountered within most of the alluvial layers, but were pervasive in the upper and lower alluvial units. The upper alluvial soils (down to about elevation 320 to 325 feet NGVD29) were observed to generally consist of medium dense, clayey or silty sand, or well graded or poorly graded sand with silt and gravel. Occasional layers of lean clay or fat clay, or clayey or well graded sandy gravels were also . The relative density of the granular soils was medium dense and the consistency of the fine-grained soils was medium stiff to stiff.

Below the upper alluvial soils, down to elevations of about 300 to 305 feet NGVD29, generally medium dense to dense soils were encountered. These soils were observed to typically consist of clayey or silty sands, or well graded or poorly graded sand with silt and gravel, or well-graded or poorly graded gravel with silt and sand. Occasional layers of lean clay or fat clay, or clayey or well graded sandy gravels were also encountered, predominately at the downstream end of the reach. The relative density of the granular soils was medium dense to dense.

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The lowest alluvial soils (from about elevations 300 to 305 feet NGVD to the top of the Puente Formation) consist predominately of sands with gravels and cobbles. These materials were generally much coarser than the upper alluvial soils, and their relative density was dense to very dense.

The bedrock encountered was of the Puente Formation (likely the Yorba Member), which was generally observed to consist of fine-grained sandstone to siltstone. The sandstone/siltstone observed was dark gray to gray, and the weathering varied from residual soil to moderately weathered rock. The bedrock was encountered at about elevations 260 to 275 feet NGVD29 downstream of about Station 1266+00, and elevations 275 to 285 feet NGVD29 upstream of that.

3.4 GROUNDWATER

Groundwater was encountered at elevations ranging from approximately 323 to 330 feet NGVD29 in the borings at the time of drilling. It is presumed that the groundwater will fluctuate with the local precipitation and river levels. Groundwater elevations measured on April 7, 2014 were about 330, 321, and 325 feet NGVD29 at wells P13-R9P5A-08, P13-R9P5A-09, P14- R9P5A-05, respectively. The anticipated groundwater levels should be updated based on continued monitoring that will be completed by USACE.

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4.0 SEISMIC AND GEOLOGIC HAZARDS

The primary seismic hazards potentially impacting the project include fault rupture, strong ground motions, soil liquefaction and slope stability. Other geologic hazards such as landslides, tsunamis, seiches, and subsidence are considered low risks for this project.

4.1 FAULT RUPTURE

The alignment is not located within any currently designated State of California Alquist-Priolo Earthquake Fault Zones (CGS, 2007). Furthermore, no active or potentially active faults are known to exist within the project improvement areas (CGS, 2010). As a result, the potential for fault rupture to impact the project is considered low. The closest active faults to the site that are considered capable of ground rupture by the California Geological Survey (CGS) are two faults that are parts of the Elsinore fault zone complex, the Whittier section (0.4 miles to the northeast), and the Chino section (9.8 miles to the northeast). Mapped active faults in the vicinity and region are shown in Figures 3 and 4.

4.2 STRONG GROUND SHAKING

A site-specific probabilistic seismic hazard analysis (PSHA) was performed to generate probabilistic response spectra and ground motion time histories for analyses of the proposed slope protection. The PSHA was performed in accordance with the guidelines presented in ER1110-2-1806, Earthquake Design and Evaluation for Civil Works Projects (USACE, 1995).

Based on EM1110-2-1806, the Operating Basis Earthquake (OBE) is defined as the earthquake with a 50% probability of exceedance within the service life of the project. The service life of the LSAR R9P5A project is 100 years. Therefore, the OBE equates to a 144 year average return period. The Maximum Design Earthquake (MDE) was selected to be equivalent to the 950 year average return period (10% probability of exceedance in 100 years) ground motion.

The PSHA was performed using the software OpenSHA (www.opensha.org). OpenSHA is a Java-based open-source platform for ground motion hazard computations, developed by the Southern California Earthquake Center (SCEC) and the United States Geological Survey (USGS). The PSHA used the standard fault model for California developed by the State of California and the United States Geological Survey for their Probabilistic Seismic Hazard maps (Field et al., 2009). The parameters of faults most relevant to the site are listed in Table 3.

http://www.opensha.org/

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Table 3 – Fault Model Parameters

Fault Name Rake Dip

Slip Rate (mm/yr) Top Depth (km)

Bottom Depth (km)

Minimum Rupture Distance

(km)

Maximum Magnitude Median Standard

Deviation

Elsinore 180 80 5 2 0 13.3 1.8 6.6 – 7.8 Chino 150 65 1 1 0 13.4 2.7 6.6 – 7.1 S San Andreas (San Bernadino S)

180 90 16 6 0 16.4 50.2 6.6 – 8.2

San Jacinto 180 90 6 4 0 16.1 40.3 7.5 – 8.1 Puente Hills 90 25 0.7 0.4 5 13 19.9 6.6 – 7.4

The four 2008 Next Generation Attenuation (NGA) relationships of Abrahamson and Silva;

Boore and Atkinson; Campbell and Bozorgnia; and Chiou and Youngs (as summarized in Abrahamson et al., 2008); were weighted equally. The NGA relations model the ground motion using site conditions specified by Vs30 (the average shear wave velocity in the upper 30 meters), together with basin depth parameters, specified by Z1.0 and Z2.5 (the depth to shear wave velocities of 1 km/s and 2.5 km/s) to incorporate basin amplification effects. The site and basin depth parameters used for the LSAR R9P5A site are:

• Vs30 = 280 m/s (NEHRP Site Class D)

• Z1.0 = 190 m

• Z2.5 = 1990 m

The Peak Ground Acceleration (PGA) for the OBE and the MDE are presented in Table 4.

Table 4 – Peak Ground Accelerations for OBE and MDE

Design Earthquake Performance Criteria

Average Return Period

(years)

Peak Ground Acceleration

(g)

Operating Basis Earthquake (OBE) Little or No Damage 144 0.281

Maximum Design Earthquake

(MDE)

No Catastrophic Failure 950 0.606

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The spectral accelerations for the OBE and the MDE, for 5 percent damping, are presented in Table 5 and in Figure 7.

Table 5 – Spectral Accelerations for 5% Damping Period (sec)

OBE

(g)

MDE

(g)

PGA 0.281 0.606

0.01 0.281 0.606

0.02 0.284 0.616

0.03 0.299 0.653

0.05 0.339 0.734

0.075 0.413 0.877

0.1 0.486 1.013

0.15 0.593 1.208

0.2 0.633 1.284

0.25 0.637 1.312

0.3 0.626 1.307

0.4 0.570 1.248

0.5 0.516 1.161

0.75 0.397 0.935

1 0.313 0.751

1.5 0.212 0.524 2 0.154 0.385 3 0.095 0.236 4 0.066 0.164 5 0.051 0.127

7.5 0.028 0.072 10 0.016 0.042

4.3 SOIL LIQUEFACTION

Soil liquefaction is a phenomenon in which loose to medium dense, saturated, granular materials develop high pore water pressure, lose shear strength and undergo matrix rearrangement because of cyclic ground vibrations induced by earthquakes. This rearrangement and strength loss is followed by a reduction in bulk volume of the liquefied soils. The secondary effects of liquefaction can include the loss of bearing capacity below foundations, settlements in level ground, and instabilities in areas of sloping ground.

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The potential for liquefaction along the project alignment was evaluated in accordance with: a) criteria and procedures proposed by Idriss and Boulanger (2008) using the Standard Penetration Test (SPT) data collected in the test holes; and b) the procedure developed by NCEER (1998) using the CPT data with the computer program CLiq (v.1.4). The potential for liquefaction was assessed in terms of a factor of safety. This factor of safety is defined as the Cyclic Resistance Ratio (CRR, the capacity of a soil layer to resist liquefaction) divided by the Cyclic Stress Ratio (CSR, seismic demand on a soil layer) generated by the design ground motion. Soils were considered potentially liquefiable if the factor of safety against liquefaction was calculated to be less than about 1.1. The potential for liquefaction was evaluated for both the OBE and MDE PGAs of 0.281g and 0.606g, respectively. An earthquake magnitude of 7.0 was used for both earthquakes.

CSR is primarily a function of the PGA generated by the earthquake. CRR is calculated based on the in-situ measurements of the soil resistance provided by the Standard Penetration Tests (SPTs) and the driving resistance of other samplers correlated to SPT sampler blow counts measured in the test holes. In the CPT analyses, CRR is calculated based on the in-situ measurements of the cone penetration resistance and the interpreted soil behavior index.

Based on the piezometer measurements, the groundwater level was assumed to range between elevations of 323 and 330 feet NGVD29 at the downstream and upstream ends of LSAR R9P5A, respectively.

The zones for which the results of the liquefaction analyses indicated factors of safety less than

1.1 are summarized in Table 6. The potentially liquefiable soils exist predominantly between elevations 325 and 300 feet NGVD29. The effects of these potentially liquefiable soils on the stability of the riverbank are discussed in Section 4.4 of this appendix.

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Table 6 – Zones of Potentially Liquefiable Soils

Exploration For OBE For MDE

Elevation of Layer (feet NGVD29)

Factor of Safety

Elevation of Layer (feet NGVD29)

Factor of Safety

TH13-R9P5A-01 - - 304 to 298 1.22

TH13-R9P5A-02 308 to 303 1.43 308 to 303 0.66

TH13-R9P5A-03 313 to 303 0.45 313 to 303 0.21

TH13-R9P5A-04 319 to 315 1.81 319 to 315 0.84 309 to 304 0.83

TH13-R9P5A-05 - - - -

CPT-01 320 to 318 0.38 to 1.03 320 to 318 0.18 to 0.47

CPT-02

324 to 322 0.52 to 0.95

324 to 322 0.16 to 0.48 316 to 314 0.24 to 0.50

312 to 307 0.36 to 0.92 312 to 307 0.16 to 0.55

CPT-03A

324 to 322 0.36 to 0.97 324 to 322 0.17 to 0.63 316 to 314 0.54 to 0.97 316 to 314 0.25 to 0.58 310 to 308 0.36 to 0.84 310 to 308 0.16 to 0.39

CPT-04A - - - -

CPT-05A 329 to 326 0.24 to 0.66 329 to 326 0.11 to 0.30

CPT-06A - - 330 to 331 0.50

TH14-R9P5A-01 306 to 301 1.08 323 to 316 1.13 306 to 301 0.50

TH14-R9P5A-02 311 to 308 1.60 322 to 316 0.94 311 to 308 0.74

TH14-R9P5A-03 - - - -

TH14-R9P5A-04 312 to 307 0.98 312 to 307 0.45 291 to 286 1.06

TH14-R9P5A-05 314 to 309 1.05 314 to 309 0.49

TH14-R9P5A-06 - - - -

TH14-R9P5A-07 - - 322 to 316 1.3

TH14-R9P5A-08

325 to 322 325 to 322 0.46 322 to 318 1.00 to 1.48 322 to 318 0.69

TH14-R9P5A-09 320 to 314 1.06 328 to 323 0.89 323 to 318 0.80 318 to 314 0.49

TH14-R9P5A-10 317 to 310 1.21 317 to 310 0.56

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4.4 STABILITY OF SLOPES PROTECTED BY GROUTED RIP RAP

The existing slopes where grouted rip rap slope protection is planned are inclined at 1.5:1 Horizontal:Vertical (H:V) or flatter. An existing retail center and a MSE wall are located adjacent to the proposed grouted rip rap bank protection along a portion of the alignment. This portion of the alignment is underlain by fills ranging between 5 and 15 feet thick over alluvial deposits. The fill and the upper part of the alluvium are primarily comprised of medium dense granular soils.

The proposed permanent face of the grouted rip rap slope protection and the temporary back cut are planned to be inclined at 2:1 H:V. Slope stability analyses were performed at two critical slope locations identified at Stations 1235+10 and 1239+60. The design slope heights (including the design scour depth) at these locations are about 32 and 34 feet, respectively. Although the slope heights were similar at both locations, the section at Station 1235+10 was considered critical due to the existence of the MSE wall and retail center above the crest of the proposed slope facing.

Slope stability analysis was performed in general accordance with the guidelines presented in EM 1110-2-1902, Slope Stability (USACE, 2003). The recommended minimum factors of safety are summarized in Table 7.

Table 7 – Recommended Minimum Factors of Safety for Slope Stability Analyses

Condition Factor of

Safety – Static Case

Factor of Safety – Rapid

Drawdown Case

Factor of Safety – Pseudo-static and Post Liquefaction

Case Temporary Back cut 1.2 NA NA

Permanent Grouted Rip Rap Slope Face 1.5 1.1 1.1

The computer program SLOPE/W (GeoStudio, 2012) was used to perform the slope stability analyses. The subsurface profile was determined based on the geotechnical information from the USACE 2013 investigation (TH13-R9P5A-01, TH-R9P5A-10, and P-R9P5A-09). Geotechnical strength parameters for the underlying materials were selected based on available test data and engineering judgment.

The analyses were performed for the following:

a) Temporary backcut condition during construction (Figures 8 and 15).

b) Permanent static condition at full scour depth (Figures 9 and 16).

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c) Pseudo-static condition at full scour depth for OBE case (Figures 10 and 17) and MDE case (Figures 11 and 18) - a seismic coefficient equal to one-half the PGA was used

(FEMA, 2003).

d) Pseudo-static condition with no scour for MDE case (Figures 12 and 19).

e) Post-liquefaction condition at full scour depth (Figures 13 and 20) - analyses were performed assuming the medium dense alluvial deposits below the groundwater would liquefy. A residual shear strength of 800 psf in the liquefied materials was assumed, based on an average SPT blowcount of 15 blows per foot in the potentially liquefiable materials (Seed and Harder, 1990).

f) Rapid drawdown condition (Figures 14 and 21) - analyses were performed to simulate the lag of the groundwater lowering when the 100-year flood (at an elevation of 336 feet) has receded to the normal river levels.

A summary of the results of the stability analyses are provided in Table 8.

Table 8 – Summary of Slope Stability Results Analysis

Case Figure

No. Condition Calculated Factor of Safety

Recommended Minimum Factor of Safety

Station 1235+10 1 8 Temporary Back Cut 1.5 1.2 2 9 Permanent; Full Scour; Static 1.6 1.5

3 10 Permanent; Full Scour; Pseudo- Static (OBE Case) 1.2 1.1

4 11 Permanent; Full Scour; Pseudo- Static (MDE Case) 0.9* 1.1

5 12 Permanent; No Scour; Pseudo- Static (MDE Case) 1.0* 1.1

6 13 Post Liquefaction 1.1 1.1

7 14 Permanent; Full Scour; Static;

Rapid Drawdown 1.3 1.1

Station 1239+60 1 15 Temporary Back Cut 1.5 1.2 2 16 Permanent; Full Scour; Static 1.5 1.5

3 17 Permanent; Full Scour; Pseudo- Static (OBE Case) 1.1 1.1

4 18 Permanent; Full Scour; Pseudo- Static (MDE Case) 0.8* 1.1

5 19 Permanent; No Scour; Pseudo- Static (MDE Case) 1.1 1.1

6 20 Post Liquefaction 1.2 1.1

7 21 Permanent; Full Scour; Static;

Rapid Drawdown 1.4 1.1

Note:

*The required minimum factor of safety was not achieved.

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As shown by the asterisk in the table above, the recommended minimum factor of safety of 1.1 was not achieved for the MDE cases. Seismic Deformation analyses were completed for the MDE ground motion, as discussed in Section 4.5.

4.5 SEISMIC DEFORMATION ANALYSES

Due to calculated lower factor of safety (than recommended minimum) for the MDE ground motions, analyses for seismic deformation of the rip rap protected slopes at Stations 1235+10 and 1239+60 were conducted. These analyses were performed using the probabilistic methodology prescribed by Bray and Travasarou (2007). Yield coefficients (ky) were determined to be 0.25 and 0.20 for Stations 1235+10 and 1239+60, respectively, using the Slope/W program. An earthquake moment magnitude of 7.0 was used in the analyses (CGS, 2002).

The median (50% probability) seismic deformations were estimated to be 4.0 and 5.5 inches at Stations 1235+10 and 1239+60, respectively. These deformations should be acceptable considering the rarity of an MDE earthquake event occurring in conjunction with full scour, which is estimated to be a result of multiple 100-year flood events occurring over the course of time.

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5.0 GEOTECHNICAL DESIGN RECOMMENDATIONS

The following sections provide geotechnical recommendations for design of the bank protection, associated earthwork, and extension of the side drains.

5.1 EARTHWORK

Earthwork will be required for the construction of the grouted rip rap slope protection and installation of the side drains. The earthwork will primarily consist of temporary excavations and backfilling, but some fills may be required to raise the bank in the areas of the grouted rip rap protection. Earthwork should be performed in accordance with the geotechnical recommendations of this report, the project specifications, and California Occupational Safety and Health Administration (OSHA) safety requirements.

5.1.1 Site Preparation

Prior to earthwork, the areas to be excavated to receive fill or to receive stockpiled materials, should be cleared and stripped of all topsoil, organic material, vegetation, abandoned utilities, and any other deleterious material or debris. Cleared and grubbed material should be properly disposed off-site. The existing rip rap will need to be removed in areas that will be protected with the new rip rap. Suitable rip rap could be reused for the project.

All areas that are to receive new fills, and the final grades in all cut areas, should be proof-rolled with heavily loaded rubber-tired equipment to locate any loose or soft areas and observed by a Geotechnical Engineer or field designate. Any loose or soft areas identified should be overexcavated and replaced with properly compacted fill soils as directed by the Geotechnical Engineer. The surface within areas to receive new fills should then be scarified to a depth of 8 inches, moisture conditioned as necessary, and compacted prior to fill placement.

5.1.2 Fill and Backfill

All fill and backfill should be compacted to a minimum relative compaction of 90 percent at a moisture content that is above the optimum moisture content. Relative compaction is defined as the ratio of the in-place dry unit weight to the maximum dry unit weight as determined by ASTM D1557. Fill soil should be placed in loose lifts no thicker than eight inches. A Geotechnical Engineer or field designate should observe all earthwork operations and perform in-place density tests.

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All engineered fill placed at the site should meet the criteria listed below:

• No oversize materials greater than 6 inches in maximum dimension.

• An Expansion Index (EI) less than 30 and a Plasticity Index less than 25.

• A relatively well-graded particle size distribution with at least 60 percent (by weight) passing a 1-inch sieve and fines content (percent, by weight, passing the No. 200 sieve) not exceeding 30 percent.

On-site soils meeting these criteria can be reused as compacted fill.

5.2 SHEET PILE WALLS

Anchored sheet pile walls are proposed to provide slope protection between Stations 1244+57 and 1274+53. It is understood that the sheet pile walls will be designed using the “Computer Program for the Design and Analysis of Sheet Pile Walls Using Classical Methods (CWALSHT)” (USACE, 1990) and “Computer Program for Winkler Soil-Structure Interaction Analysis of Sheet-Pile Walls (CWALSSI)” (USACE, 1994).

5.2.1 Geotechnical Parameters

Geotechnical engineering parameters, including lateral earth pressure coefficients were developed for use in the sheet pile wall design based on field explorations, laboratory testing and engineering judgment. Our estimation of the soil strength parameters was based on the laboratory direct shear test data and engineering judgment. A summary of the laboratory direct shear test data is presented in Table 9.

Our recommended geotechnical input parameters for the design of sheet pile walls are summarized in Table 10. The required input parameter “soil-stiffness coefficient sa/p” in the CWALSSI program is the slope of the soil pressure-displacement curves and is interpreted as “horizontal subgrade constant, lh” (Terzaghi, 1955).