C8006_Geotechnical_Report_May_2014.pdf

REVISED GEOTECHNICAL ENGINEERING REPORT

U.S. COAST GUARD CUTTER DILIGENCE MOORINGS

WILMINGTON, NORTH CAROLINA

S&ME Project No. 1306-14-009

Prepared For:

Collins Engineers, Inc.

10201 Fairfax Boulevard, Suite 530

Fairfax, VA 22030

Prepared By:

3006 Hall Waters Drive, Suite 100 Wilmington, NC 28405

NC PE Firm License No. F-0176

May 6, 2014

Revised Geotechnical Engineering Report S&ME Project No. 1306-14-009 USCG Cutter Diligence Moorings, Wilmington, NC May 6, 2014 i

TABLE OF CONTENTS

1 PROJECT BACKGROUND

1.1 INFORMATION SOURCES

1.2 SITE DESCRIPTION

1.3 PROJECT DESCRIPTION

1.4 ASSUMPTIONS

2 FIELD EXPLORATION PROGRAM

2.1 PREVIOUS SOIL TEST BORINGS BY OTHERS

2.2 TEST PITS

2.3 SOIL TEST BORINGS

3 LABORATORY TESTING

4 SUBSURFACE CONDITIONS

4.1 AREA GEOLOGY

4.2 TEST PITS

4.3 BORINGS

5 CONCLUSIONS AND RECOMMENDATIONS

5.1 SITE PREPARATION

5.2 EXCAVATION CONSIDERATIONS

Temporary Sloping and Shoring..............................................................................................65.2.1 Construction Dewatering ........................................................................................................75.2.2 Excavation Bottom Stability ....................................................................................................95.2.3

5.3 MANHOLE AND PIPE BEARING CONDITIONS

5.4 STRUCTURAL BACKFILL

Placement and Compaction.....................................................................................................95.4.1 Use of On-Site Soil as Structural Fill and Backfill................................................................105.4.2

5.5 SEISMIC DESIGN

Seismic Site Class ..................................................................................................................115.5.1 Liquefaction Evaluation ........................................................................................................115.5.2 Liquefaction Mitigation .........................................................................................................125.5.3

5.6 BULKHEAD

5.7 DEADMAN ANCHOR SYSTEM

5.8 PAVEMENTS AND SURFACE HARDSCAPING

6 LIMITATIONS OF GEOTECHNICAL REPORT

IMPORTANT INFORMATION ABOUT YOUR GEOTECHNICAL

ENGINEERING REPORT

FIGURES

APPENDIX A – BORING LOGS BY OTHERS

APPENDIX B – S&ME BORINGS LOGS/TEST PITS

APPENDIX C – LABORATORY DATA

1 PROJECT BACKGROUND

1.1 Information Sources

This report is based on the following sources of information:

Ms. Elizabeth Burkhart of Collins Engineers.

Mr. Andrew Stark, formerly with Collins Engineers.

Mr. Nate Bipes of Inland Construction and Engineering.

Preliminary topographic survey by McKim & Creed provided on April 15, 2014.

Scope of Work for Architectural-Engineering Design Services for Repair of Bulkhead at

U.S. Coast Guard Cutter Diligence Moorings, Wilmington, NC, dated December 2013, prepared by Department of Homeland Security, United States Coast Guard, Engineering Services Division.

2007 Inspection Report of the facility by Collins Engineers.

Original bulkhead design drawings dated 1964 and facility repair design drawings dated

1968, 1970, 1971, 1978, 1995, 1999, and 2005 provided by Collins Engineers.

S&ME’s past experience during previous site exploration and preliminary design phases for this project1,2,3.

1.2 Site Description

The site is between Water Street and the Cape Fear River in downtown Wilmington, North Carolina, at the approximate location shown on Figure 1. The waterfront portion of the facility consists of a steel sheetpile bulkhead with a concrete cap and a timber fender system. The northern end of the bulkhead retains a 23-feet-wide by 476-feet-long asphalt-covered wharf which services the U.S. Coast Guard Cutter (CGC) Diligence. The wharf holds the hardware used to moor the vessel and the utilities for the vessel when it is in port. The wharf is accessed by two concrete ramps located on the north and south ends of the bulkhead. The southern end of the bulkhead abuts a City of Wilmington park with asphalt pavers and planters. The sheetpile wall anchor system consists of 35- to 66-feet-long steel tie rods connected to pile-supported reinforced concrete continuous deadman. Sheet 4 of 13 of the 1964 design drawings shows three separate sections of continuous deadman along the length of the bulkhead. The piles supporting the continuous deadman anchors are shown on the original drawings as steel H-piles. The mooring cleat and bollards landward of the sheetpile wall are connected to reinforced concrete pile caps on timber piles.

1Geotechnical Engineering Report, Existing Bulkhead at U.S. Coast Guard Cutter Diligence Moorings, Wilmington, North Carolina, S&ME Project No. 1061-12-201, prepared for Collins Engineers, prepared by S&ME, dated May 11, 2012.

2 Geotechnical Engineering Report, Existing Bulkhead at U.S. Coast Guard Cutter Diligence Moorings, Stations 4+75 to 7+50, Wilmington, North Carolina, S&ME Project No. 1061-12-201A, prepared for Collins Engineers, prepared by S&ME, dated August 31, 2012.

3 Field Report of Test Pit Observations, USCG Cutter Diligence Moorings Bulkhead, dated August 6, 2013, prepared by S&ME.

1.3 Project Description

We understand that the U.S. Coast Guard has requested final design services for repair of the waterfront steel sheetpile bulkhead (approximately 750 linear feet), adjacent pavement, and waterfront accessories at the CGC Diligence Moorings. The waterfront facilities were constructed around 1965. General settlement of the adjacent pavement area, settlement of adjacent cleats/bitts and utility pits, sinkholes in the pavement along the bulkhead line, and pavement displacement above the continuous deadman anchors have been ongoing issues since that time.

Test pits and non-destructive testing in 2013 by Collins Engineers indicated that the tieback rods are in relatively good condition, but that the connections of the tieback rods with the steel sheeting have failed in some areas, resulting in excessive sheetpile deflections. Also, holes were discovered in the sheetpiling is some areas below the water line, with the loss of soil through the holes the likely cause of the sinkholes behind the bulkhead. In general, Collins Engineers recommended repairs to the bulkhead system be performed due to a combination of high stress, deadman deflection, corrosion, and fatigue.

The anticipated final repairs will likely require excavations up to approximately 10 feet deep on the river side of the continuous deadman anchors, and immediately behind the sheetpile wall to repair/rehabilitate the anchor rod connections and holes in the sheetpiling. A concrete facing will be installed on the water side of the sheetpile bulkhead, and additional piles will be installed to stabilize the continuous deadman anchors. We understand that Collins Engineers estimates that approximately 20 kips of additional lateral load resisting capacity is required for each of the existing anchor rod locations.

Also, final repairs will include extensive replacement of underground utilities at the site. These include stormwater pipe and structures, sanitary sewer, water, and electrical lines. The drawings provided to us show existing stormwater and sanitary sewer pipes and structures bearing approximately 3 to 6 feet below existing grade.

1.4 Assumptions

We assume that:

New underground utility and stormwater pipes and structures will bear less than 10 feet below the existing ground surface.

The soil and groundwater at the subject site are not contaminated

We have made these assumptions based on the project information provided to us and our experience with similar projects in the vicinity. The conclusions and recommendations given in this report are based on the background information provided to us and on our assumptions.

2 FIELD EXPLORATION PROGRAM

2.1 Previous Soil Test Borings by Others

Sheet 5 of 13 of the original project design drawings dated December 30, 1964, shows the logs and locations of 10 soil test borings taken from a Froehling & Robertson (F&R) report dated December 24, 1963. The approximate locations of these borings are shown on Figure 2 of this report, and the F&R boring logs are attached in Appendix A.

RFTS reported the results of eight soil test borings drilled along Water Street near and adjacent to the subject project site in March of 2013. These borings were drilled for the Water Street Street Scaping project planned by the City of Wilmington. The approximate locations of some of these borings are shown on Figure 2 of this report, and the applicable boring logs (B-3 through B-8) are attached in Appendix A.

2.2 Test Pits

In July and August of 2013, Inland Construction and Engineering (ICE) excavated and dewatered five test pits at the site for the purpose of observing, evaluating, and testing portions of the existing sheetpile wall, waler, tie rods, and continuous deadman. S&ME personnel visited the site on August 6, 2013, to observe the test pit excavations and site exploration activities by ICE and Collins Engineers. A field report describing S&ME’s test pit observations is attached in Appendix B.

2.3 Soil Test Borings

S&ME drillers and our subcontract driller advanced a total 19 soil test borings (B-1 through B- 5, B-2A, C-1 through C-5, and D-1 through D-8) to depths of 10 to 100 feet below the existing ground surface in 2012 and 2014. S&ME personnel located the borings in the field by measuring distances from the existing site features. The approximate boring locations are shown in Figure 2.

The soil test borings were drilled using wash boring drilling procedures. Within the borings, samples of subsurface soils were generally taken at 2.5-foot intervals within the upper 10 feet, and at 5-foot intervals thereafter using a split-spoon sampler. Within the shallow borings (C-1 through C-5), continuous samples of subsurface soils were generally taken using a split-spoon sampler. Standard penetration testing (SPT) was performed in conjunction with split-spoon sampling in general accordance with ASTM D 1586.

S&ME’s subcontract driller attempted to obtain undisturbed Shelby tube samples of the organic clay stratum from Boring B-2A. Two of the tubes were damaged and samples were not recovered due to debris in the upper fill and wood in the organic silt. Three of the attempts recovered enough material for laboratory testing (ST-3, ST-4, and ST-5).

Representative portions of the split-spoon samples were returned to our laboratory for visual-manual classification in general accordance with the Unified Soil Classification System.

Boring logs containing soil descriptions, SPT N-values, and drilling observations are included with this report in Appendix B.

The ground surface elevations shown on the attached boring logs, cross sections, and profiles were estimated from our site observations and from the spot elevations and design elevations shown on the drawings provided to us, relative to mean sea level.

3 LABORATORY TESTING

S&ME performed moisture content tests, grain-size tests, Atterberg limits tests, and organic content tests on representative samples obtained from the borings. We also performed consolidation and triaxial shear strength tests on the recovered undisturbed Shelby tube samples. These tests were performed to confirm visual soil classifications and estimate the engineering properties of the soils tested. Testing was performed in general accordance with applicable ASTM standards. Test results are attached in Appendix C.

4 SUBSURFACE CONDITIONS

4.1 Area Geology

The site is located within the Coastal Plain Physiographic Province of North Carolina. The Coastal Plain Province is typically characterized by marine, alluvial, and eolian sediments that were deposited during periods of fluctuating sea levels and moving shorelines. The soils and basal formations in the North Carolina Coastal Plain Physiographic Province are typical of those laid down in a shallow sloping sea bottom; interbedded sands and clays with irregular deposits of shells and layers of limestone and cemented sands. Alluvial sands, silts, and clays are typically present near rivers and creeks. Deposits of peat, organic silt, and organic clay are also typically present in or near current or former tidal marsh areas in the outer portion of the Coastal Plain.

According to the 1985 Geologic Map of North Carolina, the site is underlain by the Peedee Formation. This Cretaceous formation consists of greenish gray to olive black, massive, glauconitic sand, clayey sand, and clay which is locally fossiliferous and calcareous. Patches of sandy molluscan-mold limestone can be present in the upper part of this formation.

4.2 Test Pits

Test Pit No. 1, which was observed at approximately 8 feet deep on August 6, 2013, encountered sandy fill with some debris. Test Pit No. 4, which was observed at approximately 6 feet deep on August 6, 2013, also encountered sandy fill with some debris, old pipe, and an old concrete-encased conduit. Test Pit No. 3 was not being dewatered and was filled with water up to about 3 feet below the bulkhead wall cap. It also encountered sandy fill with some debris and old pipe.

S&ME did not observe Test Pit Nos. 2 and 5, as they had already been excavated and backfilled as of August 6, 2013.

According to Mr. Nate Bipes of ICE, ICE used well points, in addition to a 4 inch centrifugal diesel sump pump, to dewater Test Pit Nos. 1 and 2. Also, he stated that the evaluation work by

ICE in Test Pit Nos. 3 through 5 was performed at low tide due to seepage through gaps and holes the sheetpile wall.

4.3 Borings

The borings encountered asphalt, asphalt pavers, brick pavers, and topsoil with grass rootmat at the ground surface. Below these surficial materials, five basic soil strata were encountered within the maximum boring termination depth of 100 feet. The strata are briefly described in the following table.

Table 1 – Generalized Subsurface Strata

Stratum Average

Depth (Feet) Average Relative Density or Consistency and Description

N-avg (bpf)

From To

1 0 22 FILL: Medium Dense Sand, with zones of Debris Laden Fill

(GP, SW, SP, SP-SM, SM, SC)

2* 22 24 Loose Sand

(SM, SC)

3 24 38 Very Soft Organic Silt /Clay

(OH )

4 38 49 Medium Dense Sand

(GP, SP, SW-SM, SM, SC)

5 49 100

PEEDEE FORMATION:

Very Dense Sand with Limestone Lenses (SP, SM)

Notes: ND = Not Determined N = Standard Penetration Test blow count Bpf = Blows per foot Avg = average

* = Not uniformly present across the site.

Stratum 2 not encountered in Borings B-3, B-4, B-5, D-2, D-3, and D-4.

Obstructions encountered within the fill in the borings and observed by S&ME personnel during previous excavations on this site and nearby sites include concrete slabs and footings, sections of brick masonry, wood pieces (possibly mats), cobbles, concrete utility conduits, concrete pipes, steel rails, and metal pipes.

Please note that Boring D-5 did not encounter very soft organic silt/clay (OH) within the Stratum 3 depth range, but instead encountered very stiff to hard clay (CL) from depths of 22 to 32 feet.

Water levels were measured in the soil test borings. The water levels varied from depths of approximately 2 to 8 feet below the existing ground surface, with an average depth of 4 feet.

These depths correspond to elevations from -2 to +3 feet and an average elevation of 0 (mean sea level).

Groundwater levels have been measured or inferred in the explorations at the times and under the conditions stated on the logs in this report. Changes in the groundwater conditions and depths may occur due to seasonal variations in rainfall, evaporation, construction activity, surface water runoff, and other site specific factors, such as tides.

Subsurface cross sections are shown in Figures 3 through 7. The strata interfaces are assumed to extend out in front of the bulkhead under the Cape Fear River. The locations and elevations of the existing site features (e.g., sheetpile bulkhead, mudline, tie rods, deadmen, and H-piles) shown on the cross sections were estimated from drawings provided to us.

Subsurface profiles showing the subsurface conditions across the site are given on Figures 8 and 9 in the Appendix. The profiles represent an estimate of how the soil strata vary across the site generally in the north-south direction parallel to the Cape Fear River.

5 CONCLUSIONS AND RECOMMENDATIONS

5.1 Site Preparation

Site preparation should be initiated by clearing and stripping the proposed construction areas of any asphalt, concrete, curb, trees, shrubs, other plants, topsoil, roots, organics, and other unsuitable material. In the existing grassed areas, we anticipate an average stripping depth of 6 inches to remove the topsoil and grass rootmat.

5.2 Excavation Considerations

We assume that the planned construction (new underground utilities, new underground stormwater system, repairs to sheetpile bulkhead, and additional anchorage of the continuous deadman ) will require temporary construction excavations up to 10 feet deep. Temporary construction excavations will require temporary sloping and/or shoring (if more than 4 feet deep or below the subsurface water level that require entrance by personnel). Also, temporary construction excavations below the subsurface water level should be dewatered.

Temporary Sloping and Shoring5.2.1

Temporary construction slopes and excavation shoring should comply with local, state, and federal governing regulations, including OSHA (29 CFR Part 1926) excavation trench safety standards. Temporary excavations should be cut to a stable slope or the excavations should be temporarily braced, depending upon the excavation depth, nearby site features, and encountered subsurface conditions. The Contractor’s Competent Person will need to determine the soil type(s) encountered in excavations and the appropriate excavation slope or bracing.

Temporary slopes may need to be designed using analytical methods, as the standard OSHA temporary slope angles for different types of soils may not be applicable for this site. For example, temporary excavations sloped at 1.5H:1V for Type C soils will typically not be stable if groundwater inflow is not adequately controlled and/or removed, or if layers of very soft to soft clay or peat are prevalent in the excavation sideslopes.

Stockpiles should be placed well away from the top edge of the excavations, and their heights should be controlled so they do not surcharge the sides of the excavations. We recommend that stockpiles be set back a minimum horizontal distance from the top edge of the excavation equal to the maximum depth of the excavation.

Care should be taken not to undermine the foundations of any existing structures or utilities to remain in place.

The following soil properties can be used for design of temporary slopes or excavation shoring systems.

Table 2 – Recommended Excavation Slope/Shoring Design Parameters

Depth (feet)

Soil Description Total Unit

Weight (pcf)

Undrained Friction Angle

(degrees)

Undrained Cohesion

(psf)

Drained Friction Angle

(degrees)

Drained Cohesion

(psf)

0 - 22 FILL: Medium Dense SAND, with Zones of

Debris Laden Fill 120 30 0 30 0

Based on our site observations during the test pit excavations in 2013 and the groundwater depths measured in the borings, we recommend that temporary excavations be designed for groundwater at a typical high tide level of the Cape Fear River (approximately elevation +2 feet).

The responsibility for excavation safety and stability of temporary construction slopes should lie solely with the contractor. This information is provided only as a service and under no circumstance should S&ME be assumed to be responsible for construction site safety.

Construction Dewatering5.2.2

Potential groundwater control methods for the temporary construction excavations include open pumping, predrainage dewatering, and cutoff/exclusion.

Open pumping utilizes submersible sump pumps in pits or trenches dug below the bottom of the excavation and backfilled with No. 57 stone. Submersible sump pumps placed in perforated or slotted plastic pipe installed in angled pits backfilled with No. 57 stone below the bottom of the excavation, called “stingers”, are another type of open pumping. Pumping can be performed on an intermittent basis to remove water from the construction excavations, but should be continuous (24 hours a day) to maintain excavation bottom stability.

Predrainage dewatering requires the installation of temporary wells and/or wellpoints within or around the excavations, with continuous (24 hours a day) pumping both before excavation and while the excavations are in use to lower the groundwater adequately beneath the entire excavation areas.

Cutoff or exclusion involves the installation of sheetpiling completely around the excavations and open pumping from inside the sheeting during excavation. Predrainage dewatering outside the sheeting can be implemented to reduce groundwater pressure on the sheetpiling and potential leakage of the sheeting at joints.

Groundwater should be maintained a minimum of 3 feet below the bottoms of the temporary excavations throughout construction to maintain bottom stability, as measured in PVC standpipe piezometers screened 5 feet below the bottoms of the excavations and backfilled with clean sand (SW, SP). The contractor should install temporary piezometers on a spacing of approximately 50 feet in large, open site excavations and every 200 linear feet along trench excavations below the water table. The piezometer water levels should be measured to evaluate the effectiveness of the construction dewatering.

The dewatering system designer should consider the influence of dewatering on any nearby structures. Dewatering can cause settlement in the following ways4:

By removing fines from the soil because of improperly constructed wells or wellpoints.

By pumping from open, unbraced excavations in sands which can result in boils, piping of soil, or sloughing of slopes.

From consolidation of compressive silts, clays, or very loose sands due to an increase in effective stress.

Since the borings encountered very soft to soft organic silt/clay, there is a concern that dewatering could induce settlement due to consolidation of the foundation soils from a stress increase during dewatering. This should be evaluated by the dewatering system designer and mitigated by the contractor, as necessary. Our experience with the local loose sands is that typical dewatering methods for similar projects do not result in settlement of nearby structures.

However, if pumping from open excavations is performed improperly or well points/wells are not properly screened, loss of ground and subsidence can occur with the local sands.

Preferential horizontal permeability at the contact between any sand and clay layers can result in seeps in construction excavations at these contacts due to perching of the groundwater within the sand layer above the clay stratum. Based on the soils encountered in the borings and the maximum assumed construction excavation depth of 10 feet, we do not anticipate this condition for this project, since the excavations will be in the fill and loose sands, and will not extend down into the organic silt/clay.

Based on our site observations and the groundwater depths measured in the borings, we recommend that the construction dewatering system be designed for groundwater at a typical high tide level of the Cape Fear River (approximately elevation +2 feet mean sea level).

Based on our experience with similar subsurface conditions and similar projects in coastal North Carolina, we anticipate that predrainage dewatering and open pumping will be necessary for temporary foundation, utility, and/or stormwater pipe/structure construction excavations for

4 Powers, J.P. (1981), Construction Dewatering – A Guide to Theory and Practice, John Wiley & Sons, New York.

this project below the water table. Cutoff with temporary sheetpiling, where practical, may also be beneficial to the dewatering efforts. However, due to subsurface obstructions, including existing anchor rods, temporary sheetpiling is unlikely to result in complete cutoff of groundwater inflow into temporary construction excavations.

The responsibility for dewatering of construction excavations should lie solely with the contractor. This information is provided only as a service and under no circumstance should S&ME be assumed to be responsible for the effectiveness of the construction dewatering method(s) selected by the contractor.

Excavation Bottom Stability5.2.3

Bottom stability will be very important for construction at this site due to the anticipated excavation depths below the existing ground surface and below the groundwater level.

Excavation bottom stability will be dependent on the soil type at and below the bottom of the excavation and the effectiveness of the contractor's shoring and dewatering systems. For example, use of trench boxes and sump pumps often results in running, unstable sands at the pipe and structure bearing elevations. Use of wells and/or wellpoints to maintain the groundwater level at least 3 feet below the pipe and structure bearing elevations, along with sloped or sheeted excavations, can prevent this problem from occurring in sands which would otherwise be stable.

5.3 Manhole and Pipe Bearing Conditions

Where new manholes and pipe bear more than 3 feet below the exiting ground surface, they should bear on No. 57 stone due to zones of very loose sand and debris laden fill in the Stratum 1 fill, in addition to the potential for variations in subsurface conditions along the alignments.

The manholes should be supported on 12 inches of No. 57 stone and the pipe should be supported on 6 inches of No. 57 stone.

5.4 Structural Backfill

Placement and Compaction5.4.1

Structural backfill should be placed to backfill the temporary construction excavations for the new underground utilities, stormwater pipe and structures, and repairs to the sheetpile bulkhead and continuous deadman anchors. We recommend that a soil having a Unified Soil Classification of GW, GP, GP-GM, GP-GC, GW-GM, GW-GC, GM, SW, SP, SP-SM, SP-SC, SW-SM, SW-SC, or SM be used. The satisfactory or unsatisfactory definition of the structural fill and structural backfill materials should be based on soil classification determined by grain-size distribution and plasticity, not by moisture content. Relatively dry or wet soils meeting the recommended classifications can be used as structural fill and structural backfill, but may require wetting or drying to achieve the recommended moisture content and compaction.

The structural backfill should also be generally free of organics (less than 1% by weight per ASTM D-2974) and free of deleterious materials, such as wood. Pieces of inert debris, such as brick and concrete, should be limited to no more than 3 inches in maximum dimension, and should be blended with soil to help prevent the formation of voids.

Structural backfill should be placed in 12-inch maximum loose lifts in open site areas (more than 10 feet wide and compacted with heavier equipment) and 8-inch maximum loose lifts in trench areas (less than 10 feet wide and compacted with hand operated equipment). Structural backfill should be compacted to at least 95 percent of the standard Proctor maximum dry unit weight (ASTM D-698) more than 12 inches below finish subgrade elevations. The upper 12 inches of structural backfill should be compacted to at least 98 percent of the material's standard Proctor maximum dry unit weight. The moisture content of the material should be within +/- 3% of the material’s standard Proctor optimum moisture content for most types of satisfactory soils. The acceptable moisture content range can be increased to +/- 5% for satisfactory sands and gravels with less than 35 percent fines, provided the recommended compaction is still achieved.

Even though silty sands (SM) are satisfactory for use as compacted structural backfill, they are moisture sensitive and can be difficult to adequately dry and compact, especially during the wetter, cooler times of the year. Even if the specified compaction is achieved, the fill surface must be stable prior to placement of overlying lifts of fill, which can be difficult to achieve for these moisture-sensitive soils.

To confirm that the specified degree of compaction is being obtained, field compaction testing should be performed in each fill lift by the geotechnical engineer’s representative. We recommend that compaction tests be performed at a minimum frequency of one test per 500 square feet per lift open site excavations and one test per 50 linear feet per lift of utility trench backfill in structural areas.

Use of On-Site Soil as Structural Fill and Backfill5.4.2

The on-site soils meeting the classifications for recommended satisfactory structural backfill, plus meeting the restrictions on organic content and debris, may be used as structural backfill. Based on the test pits and borings, we anticipate that approximately half of the fill encountered within the anticipated excavation depths will be unsatisfactory for use as compacted structural backfill due to large debris or organic content (wood). Screening of the excavated material to remove excessively large and organic debris could be performed to increase the potential volume of re-usable on site soils.

On-site soils used as structural fill and structural backfill will require careful moisture control in order to achieve compaction and stability. Any soils excavated from below the water table will require significant drying to achieve the recommended moisture content and minimum compaction. Soils above the water table may also be relatively dry at the time of construction and require wetting to achieve the recommended moisture content and minimum compaction.

During excavation of on-site soils, the contractor should segregate satisfactory and unsatisfactory soils based on their classifications and other recommended restrictions. Drier satisfactory soils (e.g., excavated from above the water table) and wetter satisfactory soils (e.g., excavated from below the water table) should also be segregated. The wetter soils may require spreading, discing, and drying. Wetter sands with low fines content (SP, SW, SP-SM, SW-SM) excavated from the below the water table can typically be stockpiled and allowed to dry. However, during the typically cooler, wetter months of the year, it may be more economical to import satisfactory soils from off site, rather than to attempt to dry and re-use wet to saturated on-site soils.

The use of on-site satisfactory soils, use of imported off-site soils, wasting of on-site soils, and quantities required are subject to the project design grades and the contractor’s means and methods. These issues should be accounted for by the contractor during the bid phase of the project. The use of imported off-site soils to replace on-site satisfactory soils which may be excessively dry or excessively wet at the time of construction is acceptable, but if chosen by the contractor, should be included in the contractor’s pricing for the project.

5.5 Seismic Design

Seismic Site Class5.5.1

Per Section 1613 of the 2012 North Carolina Building Code, a Site Class E is recommended for design, based on the Standard Penetration Test N-values measured in the soil test borings and more than 10 feet of soil within the upper 100 feet of the subsurface profile having the following characteristics.

1. Plasticity Index PI > 20.

2. Moisture Content w > 40%;, and

3. Undrained Shear Strength Su<500 psf.

The Stratum 3 Organic Silt/Clay has these characteristics.

Liquefaction Evaluation5.5.2

Because of the presence of isolated zones of very loose to loose sands below the water table at the site, we performed a seismic liquefaction analysis based on the design earthquake (M=7.25). The design earthquake has a 2 percent probability of exceedance in a 50 year period. This is equivalent to an earthquake that has the likelihood of occurring once every 2,475 years. Using the 2012 North Carolina Building Code (which is based on the 2009 International Building Code), the design seismic event (magnitude 7 to 7.5) has a peak ground acceleration (pga) of 0.186g at this site. This value was calculated using Seismic Site Class E, calculating the five-percent damped spectral response acceleration at short periods, SDS, using the USGS website tool for U.S. Seismic Design Maps, and dividing SDS by 2.5 per Section

1803.5.12 (2) of the 2012 North Carolina Building Code.

Liquefaction can create two potential problems: ground surface disruption (i.e. sand boils) and volumetric compression (i.e. settlement). When soils susceptible to liquefaction are located within approximately 10 feet of the surface, ground surface disruptions are likely.

For this site, there is a risk of severe liquefaction and foundation damage is likely5 (based on Liquefaction Potential Index6 > 15). Our analyses indicate that the liquefiable soils at this site are within the approximate zones shown on Figures 3 through 7. We estimate liquefaction-induced ground surface settlement of 2 to 11 inches due to the design earthquake using the Ishiari and Yoshimine method. The wide range of estimated settlement is primarily due to the variable thicknesses of liquefiable layers of sands encountered in the soil test borings. Also, since some of the liquefiable soils are within 10 feet of the ground surface, ground surface disruptions (i.e., sand boils) are likely to occur under the design earthquake event in localized areas.

Liquefaction Mitigation5.5.3

The repairs to the existing facility should be designed to meet the current seismic design code requirements, including the effects of potential liquefaction. The analysis and design of the bulkhead and anchor piles should include seismic forces, in addition to the effects of cyclic loading and potential liquefaction on the properties of the soils causing lateral earth pressures on the sheetpile bulkhead and providing anchor pile support.

5.6 Bulkhead

The bulkhead walls must be capable of resisting lateral earth pressures that will be imposed on them. We recommend that the in situ soil properties given on the subsurface cross sections (Figures 3 through 7) be used for analysis of the existing bulkhead wall and design of repairs to the bulkhead wall. The analysis and design of the bulkhead walls should also include seismic loads; surcharge loads due to vehicles, equipment, or other live loads acting at the top of the walls; and differential hydrostatic pressures between the groundwater behind the wall and the water level of the Cape Fear River at low tide.

The bulkhead should be analyzed and repairs designed using dynamic pressures resulting from the design earthquake event. The dynamic pressures acting on yielding walls are usually estimated by pseudostatic procedures. Pseudostatic acceleration values of one-third to one-half of the peak ground surface acceleration are commonly used with factors of safety of 1.0 to 1.27.

Since the bulkhead sheetpile wall can likely move sufficiently to develop minimum active earth pressures, it is assumed to be a yielding wall. Therefore, we used the Mononobe-Okabe Method to estimate a dynamic (i.e., seismic) active thrust of 4,400 pounds per foot acting horizontally 13 feet up from the mudline for this project. We recommend that Collins Engineer’s analyses of the existing bulkhead’s stability and any improvements/repairs take this seismic loading into account.

5 Toprak, S., and Holzer, T. L. (2003). “Liquefaction potential index: Field assessment.” J. Geotech. Geoenviron.

Eng., 129 (4), 315-322.

6 Iwasaki, T., Tokida, K., Tatsuoka, F., Watanabe, S., Yasuda, S., and Sato, H. (1982). Microzonation for soil liquefaction potential using simplified methods.” Proc., 3rd Int. Earthquake Microzonation Conf., US National Science Foundation, Washington, DC., 1319–1330.

7 Kramer, S.L. (1996), Geotechnical Earthquake Engineering,Prentice Hall, Upper Saddle River, New Jersey..

The wall backfill should be compacted to at least 95 percent of the standard Proctor maximum dry unit weight (ASTM D 698). When compacting the backfill, heavier self-propelled compaction equipment should not be used within 5 feet of the structure walls. We caution that operating compaction equipment directly behind the retaining structures can create lateral earth pressures far in excess of those recommended for analysis and design.

5.7 Deadman Anchor System

The soil properties given on the cross sections (Figures 3 through 7) can be used to evaluate the structural stability of the existing continuous deadman anchor system and design repairs to the system. Soil properties for the liquefiable zones of soil are also given on these figures.

We understand that Collins’ preliminary analyses indicate that the continuous deadman anchors have deflected excessively and that approximately 20 kips of additional lateral load resisting capacity is required for each of the existing anchor rod locations. Additional piles are needed to stabilize the continuous deadman anchors and provide the needed 20 kips of additional resistance. Preliminary designs incorporate H-piles (10x57 and 12x74 sections) up to 60 feet in length, some battered at 5:12 (i.e., 22.6 degrees from vertical). The pile heads will be embedded in new pile caps to be constructed on the river side of the existing continuous deadman anchors approximately 8 to 12 feet below the existing ground surface. A single battered H-pile and single vertical H-pile are shown schematically for each of the subsurface cross sections in Figures 10 through 14.

The H-piles should be driven with an impact or vibratory hammer to bear on the Peedee Formation and should be designed using the allowable loads given in the following table:

Table 3 – H-Pile Tip Depths and Allowable Design Loads

H-Pile Section

Estimated Tip Depth Below

Ground Surface (feet)

Allowable Geotechnical Design Capacity

Compression Load (kips)

Tension Load (kips)

Lateral Load (kips)

HP 10x57 50 to 60 50 10 5

HP 12x74 50 to 60 70 15 7

Pile capacity analyses were performed to estimate axial compression and axial tension load capacities of single vertical piles based on a factor of safety of 2.0. Finite difference soil-structure interaction analyses were performed using the LPILE 2012 software to estimate the allowable lateral loads which limit lateral pile head deflection to ½ inch for a free head condition and ¼ inch for a fixed head condition in the strong-axis direction of the H-piles. The allowable lateral loads are relatively low for H-piles in similar soils due to the zones of liquefiable sands at and below the pile head elevation.

Evaluation of pile group capacities was beyond the scope of S&ME’s services, but should be performed by Collins Engineers to evaluate the loading on the pile group configuration selected. Also, Collins Engineers should evaluate the ability of the pile sections to structurally resist the internal compression, tension, shear, and moments produced by axial and lateral loading.

To install the H-piles, we anticipate that pre-drilling and/or pre-excavation to remove obstructions will be required at many locations at the site. Obstructions encountered within the Stratum 1 fill in the borings and observed by S&ME personnel during previous excavations on this site and nearby sites include concrete slabs and footings, sections of brick masonry, wood pieces (possibly mats), cobbles, concrete utility conduits, concrete pipes, steel rails, and metal pipes.

If battered or vertical piles are to be installed by an impact hammer, a wave equation analysis should be performed to verify that the proposed driving system (i.e., hammer type and size) is capable of driving the piles to the desired depth and to establish the driving criteria prior to construction. Driving criteria and pile lengths should be established and verified during driving by the geotechnical engineer. We recommend that three over-length test piles be driven to establish driving criteria and pile lengths. Load testing of the test piles should be conducted during installation. We recommend that testing be performed using a Pile Driving Analyzer (PDA) under the direction of the geotechnical engineer. Past experience indicates that ultimate mobilized pile bearing capacities increase with time for steel H-piles driven into the Peedee Formation due to pile "set-up". Pile set-up results from an increase in effective soil shear strength as the excess pore water pressures generated by pile driving in the Peedee Formation dissipate with time. Therefore, we recommend that restrikes and PDA testing of the test piles be performed at least 24 hours after initial installation.

If battered or vertical piles are installed by a vibratory hammer, we recommend that one over-length test pile be installed to monitor penetration rates and pile lengths. We recommend static load testing be conducted at least 24 hours after the test pile has been installed to allow for pore pressure dissipation. The full deflection of the test pile should be observed and recorded. The load test should be performed in accordance with the standard testing procedures presented in ASTM D-1143 and the requirements of the North Carolina Building Code. The test pile should be loaded to 300% of the design (i.e., allowable) compression load or plunging, whichever occurs first.

If vertical piles alone (i.e., no battered piles) are installed by either an impact hammer or vibratory hammer to achieve lateral load resistance, we recommend that the project include one lateral pile load test performed in accordance with ASTM D3966 and the 2012 North Carolina Building Code. The test pile should be laterally loaded to 300% of its design load.

Pile installation operations should be observed by the geotechnical engineer to monitor that the piles are encountering expected driving resistances and lengths, and to note any damage or other concerns during installation. Only by observing the installation of the piles and recording the penetration resistance values during installation can the geotechnical engineer document that the piles have generally been installed per the project drawings, specifications, and recommendations provided in this report.

Due to the noise and vibrations generated by H-pile installation, we recommend that pre-construction surveys, vibration monitoring, and post-construction surveys be performed on nearby buildings, such as the Federal Courthouse across Water Street from the Coast Guard’s non-operational area and the multi-family, multi-story residential building across Water Street from the Coast Guard’s operational area.

5.8 Pavements and Surface Hardscaping

Even after the planned repairs to the bulkhead sheetpile wall and deadman anchor system, excessive long-term total settlement and differential settlement of new pavements and hardscaping are still anticipated at the project site due to the following factors:

Downward flow of sand by surface water infiltration and tidal fluctuation into voids in the underlying debris laden fill.

Long-term secondary consolidation settlement of the very soft to soft organic silt/clay.

Based on the consolidation test results, we estimate that primary consolidation of the organic silt/clay at the site due to the original placement of the overlying fill materials would have resulted in a total settlement of up to approximately 3 to 6 feet over a period of 2 to 4 years, assuming all the overlying fill was originally placed at the same time. Primary consolidation is complete, but secondary, long-term consolidation of the organic silt/clay will continue to occur at a rate of approximately 1 to 2 inches every 50 years.

To mitigate the potential for excessive long-term total and differential settlement we recommend that geogrid-reinforced gravel mats be constructed under the pavement/hardscape subgrade, as show in Figure 15. The pavement repairs could include the wharf area, the city park area subjected to vehicular traffic, and Water Street. Hardscapes include sidewalks, pavers, etc., in areas that will not be subjected to vehicular traffic. This approach provides the lowest construction cost than previously identified alternatives, but has the greatest risk of long-term excessive total and differential settlement compared to the other alternatives.

Also, the pavements and hardscaping are expected to settle 2 to 11 inches due to liquefaction caused by the design earthquake event. Typically, pavements and hardscapes can settle this much without resulting in catastrophic, life-threatening failure. However, significant post-earthquake repairs would be required to return the facility to full functionality. The recommended geogrid-reinforced gravel mats will decrease the potential for severe differential liquefaction-induced settlement damage to the pavement and hardscape surfaces.

6 LIMITATIONS OF GEOTECHNICAL REPORT

This report has been prepared in accordance with generally accepted geotechnical engineering practice for specific application to this project. Any wetland, environmental, or contaminant assessment efforts are beyond the scope of this geotechnical exploration; and therefore, those issues are not addressed in this report. The conclusions and recommendations contained in this report are based upon applicable standards of our practice in this geographic area at the time this report was prepared. No other warranty, expressed or implied, is made.

Conclusions and recommendations submitted in this report are based, in part, upon the data obtained from the geotechnical exploration. The nature and extent of variations between and away from the borings, soundings, and/or test pits may not become evident until construction.

If variations appear evident, then it will be necessary to re-evaluate the recommendations of this report. In the event that any changes in the grades, nature, design, or location of the proposed project are planned, the recommendations contained in this report should be reviewed and modified or confirmed in writing.

S&ME should be provided the opportunity to review the final plans and specifications in order that earthwork, foundation, and other recommendations are properly interpreted and implemented. The recommendations in this report are contingent on S&ME’s observation and monitoring of geotechnical construction activities.

Portion obtained with permission from “Important Information About Your Geotechnical Engineering Report”, ASFE, 2004 © S&ME, Inc. 2010

Important Information About Your

Geotechnical Engineering Report

Variations in subsurface conditions can be a principal cause of construction delays, cost overruns and claims.

The following information is provided to assist you in understanding and managing the risk of these variations.

Geotechnical Findings Are Professional Opinions Geotechnical engineers cannot specify material properties as other design engineers do.

Geotechnical material properties have a far broader range on a given site than any manufactured construction material, and some geotechnical material properties may change over time because of exposure to air and water, or human activity.

Site exploration identifies subsurface conditions at the time of exploration and only at the points where subsurface tests are performed or samples obtained. Geotechnical engineers review field and laboratory data and then apply their judgment to render professional opinions about site subsurface conditions. Their recommendations rely upon these professional opinions. Variations in the vertical and lateral extent of subsurface materials may be encountered during construction that significantly impact construction schedules, methods and material volumes. While higher levels of subsurface exploration can mitigate the risk of encountering unanticipated subsurface conditions, no level of subsurface exploration can eliminate this risk.

Scope of Geotechnical Services Professional geotechnical engineering judgment is required to develop a geotechnical exploration scope to obtain information necessary to support design and construction. A number of unique project factors are considered in developing the scope of geotechnical services, such as the exploration objective; the location, type, size and weight of the proposed structure; proposed site grades and improvements; the construction schedule and sequence; and the site geology.

Geotechnical engineers apply their experience with construction methods, subsurface conditions and exploration methods to develop the exploration scope. The scope of each exploration is unique based on available project and site information.

Incomplete project information or constraints on the scope of exploration increases the risk of variations in subsurface conditions not being identified and addressed in the geotechnical report.

Services Are Performed for Specific Projects Because the scope of each geotechnical exploration is unique, each geotechnical report is unique. Subsurface conditions are explored and recommendations are made for a specific project.

Subsurface information and recommendations may not be adequate for other uses. Changes in a proposed structure location, foundation loads, grades, schedule, etc. may require additional geotechnical exploration, analyses, and consultation. The geotechnical engineer should be consulted to determine if additional services are required in response to changes in proposed construction, location, loads, grades, schedule, etc.

Geo-Environmental Issues The equipment, techniques, and personnel used to perform a geo-environmental study differ significantly from those used for a geotechnical exploration. Indications of environmental contamination may be encountered incidental to performance of a geotechnical exploration but go unrecognized. Determination of the presence, type or extent of environmental contamination is beyond the scope of a geotechnical exploration.

Geotechnical Recommendations Are Not Final Recommendations are developed based on the geotechnical engineer’s understanding of the proposed construction and professional opinion of site subsurface conditions.