NM FLAP 152(1) Geotech Rpt Final.pdf

NM SR 152 BRIDGE REPLACEMENT

NM FLAP 152(1)

Final Geotechnical Report Report # NM-FX-152-18-01

Geotechnical Section Technical Services Branch

July 2018

TABLE OF CONTENTS

SECTION 1-- INTRODUCTION

1.1 SITE CONDITIONS

1.2 PROPOSED CONSTRUCTION

1.3 OBJECTIVE AND SCOPE

1.4 GEOLOGY

1.4.1 Regional Geology

1.4.2 Site Geology

1.5 GEOLOGIC HAZARDS

1.6 SEISMICITY

1.7 SEISMIC DESIGN PARAMETERS

SECTION 2-- FIELD INVESTIGATION

2.1 SUMMARY OF SUBSURFACE EXPLORATION

2.2 SUMMARY OF SUBSURFACE EXPLORATION

2.3 LABORATORY TESTING

2.4 ROCK STRENGTH EVALUATION

2.4.1 Rock Strength from Laboratory Tests

2.4.2 Generalized Strength Criterion for Intact Rock

2.4.3 Fracture Shear Strength

2.4.4 Rockmass Characterization

SECTION 3-- ANALYSIS AND RECOMMENDATIONS

3.1 SITE CHARACTERIZATION

3.2 DISCONTINUITIES AND KINEMATIC ANALYSIS

3.2.1 Proposed Abutment Foundations

3.2.2 Reinforcement Design

3.3 ABUTMENTS AND RETAINING WALL DESIGN RECOMMENDATIONS

3.3.1 Bearing Resistance for Abutments and Retaining Walls

3.3.2 Earthwork Recommendations

3.3.3 Erosion and Corrosion Protection

3.4 CUT SLOPE RECOMMENDATIONS

3.5 CONSTRUCTION CONSIDERATIONS

3.6 LIMITATIONS

SECTION 4-- REFERENCES

TABLES

Table 1: - Fault Data Table 2: - Summary of Seismic Parameters Corrected for Site Class B Table 3: - Correlation between Rock Quality and RQG Values Table 4: - Summary of Borings Table 5: - Uniaxial Compression Test Results (ASTM D7012 Method C Table 6: - Summary of Rock Strength Results from Laboratory Tests Table 7: - Mohr-Coulomb Strength Parameters for Intact Rock Based on Hoek and Brown Criterion

Table 8: - Mohr-Coulomb Strength Parameters for Fracture Shear Strength Based on Hoek and Brown Criterion Table 9:- Rock Mass Rating Table 10: - General Meaning of GSI Rating Applied to Fractured Rock Masses Table 11: - Estimated Range of GSI Table 12:- Mohr-Coulomb Strength Parameters Calculated for the Rockmass Based on Hoek and Brown Criterion Table 13:- Retaining Wall Locations and Geometry Table 14:- Estimated Soil Properties for Design Table 15:- Resistance Factors for Abutment and Wall Foundation Design Table 16:- Bearing Resistance of Overburden Materials at the Strength Limit State Table 18:- Recommended Wall Backfill Parameters for Design Table 19:- Recommended Cut Slope Ratios by Station Table 20:- Rockfall Catchment Ditch Width per Slope Height

APPENDICES

Appendix A: Project Location and Geologic Map Appendix B: Field Exploration Program Appendix C: Laboratory Test Results Appendix D: Photos Appendix E: Special Contract Requirements

SECTION 1--INTRODUCTION

This report presents the geotechnical engineering study and recommendations pertaining to the proposed bridge replacements for the NM FLAP 152(1) project. The bridges are located along New Mexico State Road 152, in Sierra County, west of Hillsboro. The first bridge (NM Bridge #1520) is approximately 5 miles west of Hillsboro and the second bridge (NM Bridge #1521) is approximately 7 miles west of Hillsboro. The Gila National Forest is immediately west of the bridges. State Route 152 is a major connector, servicing several communities, the Town of Silver City, state parks, campgrounds, and the Gila Cliff Dwellings National Monument. The route connects Interstate 25 to the east with US Highway 180 on the west. The road is currently maintained by the New Mexico Department of Transportation (NMDOT). A project location and site map are included in Appendix A.

1.1 SITE CONDITIONS

The existing 100-foot long single-span truss bridges were originally constructed in 1929 and are 19-feet wide, with a vehicle height restriction of 12.5 feet. The existing bridges and wing walls are supported on spread footing foundations bearing on competent bedrock. Each bridge currently has a major condition rating of 4 or less indicating “poor” condition and an operating rating of HS-22.

The limited width and height of the existing bridges limits vehicular access and could potentially impact emergency responders. Due to the geometric and structural obsolescence of these bridges, replacement with new structures is proposed. Photos of the existing bridges are included in Appendix D.

1.2 PROPOSED CONSTRUCTION

The proposed project consists of constructing two new bridges and the destruction of one of the two existing truss bridges. The existing western-most Bridge (#1520) is to remain in place requiring an alignment shift of State Route 152 to accommodate the new bridge. For this reason, the approach road alignments for Bridge #1520 will require substantial earthwork involving both cut and fill operations to accommodate the shift. The approach road alignment of the eastern-most Bridge (#1521) will not be substantially changed from existing, therefore cut and fill operations are anticipated to be minimal. Proposed bridges will accommodate two 12-foot travel lanes with 5-foot shoulders on each side and will be supported on spread footings bearing on bedrock. Temporary detours are anticipated as the State Route 152 must remain open during construction.

1.3 OBJECTIVE AND SCOPE

The objectives of this study are to develop geotechnical recommendations concerning bridge foundations including foundations types and geometry, geotechnical resistances at all service limit states, and anticipated settlement. In accomplishing these objectives, the scope of work included visual site reconnaissance and geological mapping, subsurface exploration, laboratory testing, correlation of available data, engineering analysis, and foundation recommendations.

Geology and Seismicity

1.4 GEOLOGY

1.4.1 Regional Geology

The bridges are located within the Mexican Highland section of the Cascade-Sierra Mountains province. The Mexican Highland is a large semi-arid plateau that extends from the Southern Rocky Mountains in the north to the Trans-Mexican Volcanic belt in the south. It is bounded by the Sierra Madre Occidental and Sierra Madre Oriental on the west and east, respectively.

The bedrock in the project area is dominated by volcanic, pyroclastic, and volcaniclastic rocks, generally of tertiary age (Hedlund, 1977). Volcanic units include quartz latite, rhyolite, and andesite flows, flow breccias, and intrusive dikes and plugs. Older sedimentary rocks of the Devonian to Cretaceous age underlie the Tertiary volcanics. The Beartooth Quartzite is of Cretaceous age and consists of white to light greenish-gray medium to coarse grained sandstone.

Underlying the Beartooth Quartzite is the Lower Permian Abo Formation, which consists of grayish-red silty mudstone with local limestone modules. The Abo Formation is underlain in turn, by the Devonian Percha Shale, a black to medium-gray shale with limestone nodules. A regional geologic map is included in Appendix A.

1.4.2 Site Geology

Bridge #1520 is mapped as underlain by Quaternary alluvium and Tertiary Mimbres Peak Rhyolite. The Mimbres Peak Rhyolite is described as light-gray to very light-gray rhyolitic lava flows and domes. Flow laminae are commonly contorted with steep dips adjacent to extrusive domes.

Bridge #1521 is mapped as underlain by Quaternary alluvium, Tertiary andesite dikes and plugs, and Permian Abo Formation. The andesite dikes and plugs are described as brownish-gray, olive-to medium-gray, and pale red-purple andesite, that weathers to dark yellowish brown. The Abo Formation is described as grayish-red silty mudstone with local limestone nodules and thin interbeds of siltstone, sandstone, and conglomerate.

1.5 GEOLOGIC HAZARDS

Geologic hazards exist from both the natural environment of the project sites and from existing and proposed construction of the roadway. The main geologic hazards that may exist within the vicinity of the project limits include: rock falls, seismic shaking, and flooding. In addition, hard bedrock is mapped near the surface in the project area and may be encountered in shallow excavations for bridge foundations.

1.6 SEISMICITY

Several faults and fault zones are located in the vicinity of the project site. The only major fault within a 40-mile range to the project location is the Caballo Fault. Other faults are more distant and summarized in Table 1 below. These faults are typically normal faults with slip-rates of 0.03 to 0.2 millimeters per year and vary in length from 19 miles, as seen in the Organ Mountains Fault, to 81 miles in the San Andres Mountain Fault. The overall trend of the faults within 40 miles of the project location is north-northeast to north-northwest with high-angle dips to the west or east.

Table 1: - Fault Data

FAULT OR FAULT ZONE

DISTANCE

FROM

PROJECT

FAULT

PARALLEL

SLIP RATE

FAULT

LENGTH AGE

(miles) (mm/year) (miles) (years)

Caballo Fault 22.44 0.14 26 <1,600,000

San Andres Mountain Fault 62.64 0.13 81 <15,000

Socorro Canyon Fault Zone 75.2 0.04 30 <130,000

Organ Mountains Fault 75.34 0.20 19 <1,600,000

La Jencia Fault 82.29 0.03 21 <750,000

Alamogordo Fault 94 0.14 75 <1,600,000

East Franklin Mountains Fault 94.04 0.13 30 <130,000

1.7 SEISMIC DESIGN PARAMETERS

Recommended seismic response parameters for the NM SR152 bridge replacement design is based on the (AASHTO) LRFD Bridge Design Specifications, latest edition, and represents horizontal peak ground acceleration (PGA) with 7 percent probability of exceedance in 75 years (approximately 1000-year return period). The 1000-year return period uniform hazard spectrum for the NM SR152 site was taken at 32.91877º N latitude and 107.6496º W longitudinal, was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short- and long- period ground acceleration (Ss and S1 respectively) and corrected for the soil profile at the end project.

Ground motions caused by earthquakes are influenced not only by the distance from the fault planes, but also by the geology and soils found at the site. Subsurface profiles similar to the ones within the project, with soil layers overlaying bedrock that significantly differ in stiffness and density, will have amplified ground motions and a resonant period governed by the layer thickness and shear wave velocity of the materials. The damage potential of strong ground motions with respect to a structure is typically affected by the period of the strong earthquake motion and the resonant period of both the soil and the structure. Although no actual shear wave velocity measurements were taken, based on the material encountered within the project area, and that bedrock outcrops at the surface, the average time-weighted shear wave velocity for the top 100 feet (VS100) of subsurface materials was estimated between 2,500 and 5,000 feet per second.

Therefore, the site is classified as Class B according to the site class definitions specified in Table 3.10.3.1-1 of AASHTO.

The recommended spectral acceleration coefficient values for probabilistic design with a return period of 1000 years were calculated using the program provided with the AASHTO LRFD Bridge Design Manual developed by the USGS (2008) entitled “Seismic Design Parameters”, version

2.10 and are summarized in Table 2, Summary of Seismic Parameters Corrected for Site Class B.

Table 2: - Summary of Seismic Parameters Corrected for Site Class B.

SEISMIC PARAMETERS

Horizontal Peak Ground Acceleration, (As) 0.070g

Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDs) 0.162g

Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.051g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.00

Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.00

Site Factor at Long-Period Range of Acceleration Spectrum, (Fv) 1.00

Based on the long acceleration coefficient SD1 value of 0.051 calculated as Fv.S1, the alignment is assigned to seismic hazard Zone 1 according to Table 3.10.6-1 in AASHTO. Seismic hazard zones reflect the variation in seismic risk in different regions needing different requirements for design as depicted in Table 4.7.4.3.1-1 in AASHTO.

SECTION 2--FIELD INVESTIGATION

2.1 SUMMARY OF SUBSURFACE EXPLORATION

Geotechnical field investigations were conducted by CFLHD geotechnical personnel in May and June 2017, for the NM 1520 and 1521 replacement bridges to characterize subsurface conditions.

The field program consisted of drilling test holes using percussion hammer and HQ coring techniques near each of the existing abutment locations. Public meetings and project design team meetings following the completion of the subsurface exploration, indicated a preference to keep the road open during construction, and to preserve Bridge #1520. Therefore, the project design team elected to shift the road alignment near Bridge #1520 approximately 40 feet east. Since the subsurface investigations were completed prior to this decision, the subsurface conditions at the proposed abutments of the replacement Bridge #1520 are less well defined, and conservative values will be assumed based on the results of the nearby exploration locations.

Four borings, one at each existing abutment, The borings, 4 in total were cored by Resilient Drilling of Mesa, AZ.

with a CME-55, truck mounted drill rig. Percent core recovery and rock quality designation (RQD) were also determined in the field for each core run to provide a qualitative basis of the rockmass competency and structural integrity conditions. Table 3 provides a rock quality description based on measured RQD Values. Photos of the core samples are presented in Appendix D.

Table 3: - Correlation between Rock Quality and RQG Values

MEASURED RQD VALUES (%) DESCRIPTION OF ROCK QUALITY

0-25 Very Poor

25-50 Poor

50-75 Fair

75-90 Good

90-100 Excellent

The borehole detailed stationing and offsets are based on the 30% plan set and are presented in Table 4 below.

Table 4: - Summary of Borings

TEST HOLE

NUMBER LOCATION DEPTH (feet) STATION OFFSET

(feet)

GROUND

ELEVATION

(feet)

BH17-01 EAST ABUTMENT

BRIDGE 1520 19 18+30 20 RT 5736

BH17-02

WEST

ABUTMENT

BRIDGE 1520

21 15+70 7 LT 5752

BH17-03 EAST ABUTMENT

BRIDGE 1521 28 25+20 20 LT 5963

BH17-04

WEST

ABUTMENT

BRIDGE 1521

28 23+80 18 RT 5964

2.2 SUMMARY OF SUBSURFACE EXPLORATION

Specific site descriptions are based on field observations, subsurface characterization information, and may vary in material type strength with distance away from the test boring locations. This variability is due to the topography and land forming process.

2.3 LABORATORY TESTING

Representative rock core samples were selected and tested in the laboratory to determine specific geotechnical strength properties. The unconfined compressive strength (UCS) of the rock was tested using the Uniaxial Compression Test (ASTM D 7012 Method C). A summary of the test results is presented in Table 5. The test results are presented in Appendix C.

Table 5: - Uniaxial Compression Test Results (ASTM D7012 Method C

TEST HOLE

NUMBER LOCATION

LENGTH TO

DIAMETER

RATIO

(L/D)

SAMPLE

DEPTH

(feet)

UNCONFINED

COMPRESSIVE

STRENGTH

(psi)

BH17-01 E ABUTMENT BRIDGE 1520 2.5 15.5 6,270

BH17-01 E ABUTMENT BRIDGE 1520 2.4 18.25 8,520

BH17-02 W ABUTMENT BRIDGE 1520 2.4 15.0 6,580

BH17-03 E ABUTMENT BRIDGE 1521 2.4 15.5 19,150

BH17-03 E ABUTMENT BRIDGE 1521 2.4 18.25 7,720

BH17-04 W ABUTMENT BRIDGE 1521 2.4 19.0 3,250

2.4 ROCK STRENGTH EVALUATION

Structures supported on massive bedrock are highly dependent on the overall discontinuous rockmass strength and deformation characteristics which are considerably less than those of the intact rock. The loads supported on the abutments of a beam and girder bridge are typically vertical. Therefore, the rockmass load bearing capacity is highly dependent on the rockmass strength and fracture characteristics to resist the loads applied in various directions. The rockmass is generally comprised of intact blocks of various sizes defined by intersecting fractures or discontinuities. Assessing the foundation stability in the rockmass requires reliable data on the orientation and dimension of the fractures and their shear strength to determine the resistance to sliding. The strength and deformation characteristics of the rockmass are therefore a function of the (1) intact rock strength, (2) rock discontinuity shear strength (which is required for the kinematic stability analysis), (3) rock block sizes, and (4) degree of freedom to move rock blocks. The range of the intact rock strength and the shear strength of rock fractures are determined from the laboratory test results while the block sizes are determined through surface and borehole discontinuity mapping. The rockmass strength properties rather than the intact rock properties are calculated and should be used for design purposes. Due to presence of discontinuities such as joints, bedding planes, and faults, in situ rockmass properties are always significantly lower than the intact rock properties. The following sections will determine the rockmass properties based on field and laboratory data collected during the investigations at this site.

2.4.1 Rock Strength from Laboratory Tests

The laboratory test results on core samples, presented in the previous section, were used to characterize the intact rock strength, the fracture shear strength, and the splitting tensile strength for the bedrock. The laboratory tests included uniaxial compressive strength (UCS) tests. The test results were combined based on the failure mode of each core sample (shear, fracture, bedding, substance, and combination). The overall strength of the predominant rock types impacting the foundation design of the proposed NM1520 and NM1521 bridges were analyzed. The abutments of both bridges will be supported on flow breccia and andesite. Based on the geological evaluations, the laboratory test results, and surface geological mapping, these units varied in strength between approximately 3,500 to 19,000 psi.

Six tests were performed on flow breccia and andesite core samples to obtain the rock UCS properties. In all tests, the rock failed along a fracture or a shear plane. The statistical analysis of the test results is summarized in Table 6. Based on the rock strength results listed above, and according to rock strength grades specified by the International Society for Rock Mechanics (ISRM, 1981), both the flow breccia and andesite classify as "Strong Rock– R4".

Table 6: - Summary of Rock Strength Results from Laboratory Tests

TYPE NO. OF

SAMPLES

MEAN UCS

VALUE

(psi)*

STANDARD

DEVIATION

(psi)

MINIMUM UCS

MAXIMUM UCS

Intact Rock Unconfined

6 8,582 5,481 3,250 19,150

2.4.2 Generalized Strength Criterion for Intact Rock

The empirical failure criteria developed by Hoek and Brown (1988, 1994) was used to determine the Mohr-Coulomb failure envelope for the intact rock. Hoek and Brown’s criterion for the maximum principal effective stress (𝜎𝜎1′) at failure is given by the following equation:

𝜎𝜎1′ = 𝜎𝜎3′ + 𝜎𝜎𝑐𝑐𝑐𝑐 �𝑚𝑚𝑐𝑐 ∗ 𝜎𝜎3′ 𝜎𝜎𝑐𝑐𝑐𝑐 + 1� 𝛼𝛼 where, 𝜎𝜎1′ = maximum effective principal stress 𝜎𝜎3′ = minimum effective principal stress 𝜎𝜎𝑐𝑐𝑐𝑐 uniaxial compressive strength of intact rock

The coefficients mi and 𝛼𝛼 are a function of the uniaxial compressive strength and the tensile strength of the intact rock (Sheorey, 1997; Mostyn and Douglas, 2000). Ideally, the values of mi and 𝛼𝛼 are found through a series of triaxial compression tests, but a reasonable estimate of mi is within the following limits (Mostyn and Douglas, 2000):

𝜎𝜎𝑡𝑡

− 1 < 𝑚𝑚𝑐𝑐 < 𝜎𝜎𝑡𝑡 where, 𝜎𝜎𝑐𝑐𝑐𝑐 = uniaxial compressive strength of intact rock 𝜎𝜎𝑡𝑡 = splitting tensile strength of intact rock

For intact rock strength, the value of 𝛼𝛼 is assumed to be 0.5.

Based on this method, estimates of intact rock Mohr-Coulomb strength parameters, φ ' and c', can be determined by plotting several Mohr's circles for (σ1′ , σ3′ ) pairs and the failure envelope. Table 7 shows the resulting estimated intact strength parameters using the average and minimum uniaxial compressive strength values.

Table 7: - Mohr-Coulomb Strength Parameters for Intact Rock Based on Hoek and Brown Criterion

BEDROCK TYPE σci (psi) mi α

ESTIMATED INTACT MOHR-COULOMB

STRENGTH PARAMETERS

φ' (degrees) c' (psi)

Andesite 8,582* 20 0.50 55 1,342 Andesite 3,250** 20 0.50 54 4,990

* Average test value for all samples

** Minimum test value for all samples

2.4.3 Fracture Shear Strength

Fracture shear strength is used to analyze failures through the geological structures. The rockmass strength is also a function of both intact rock strength and fracture shear strength, and is influenced by the distribution and size of the intact rock blocks and by the orientations of the persistent geological structures (fractures, faults…) present in the rockmass. The fracture shear strength generally influences the theoretical lower bound of the rockmass strength.

No direct shear tests were performed on andesite core specimens to determine the range of fracture shear strength due to the inconsistency of the measured fracture sets. Assumed values of basic friction angle were made based on intact rock using the Hoek and Brown (1988) criteria described in Section 3.4.2 above. Table 8 shows the resulting estimated fracture shear strength parameters using both the average and minimum uniaxial compressive strength values.

Table 8: - Mohr-Coulomb Strength Parameters for Fracture Shear Strength Based on Hoek and Brown Criterion

ROCK TYPE σci (psi) mi α

ESTIMATED FRACTURE MOHR-COULOMB

SHEAR STRENGTH PARAMETERS

φ' (degrees) c' (psi)

Andesite 8,582* 7 0.50 48 82 Andesite 3,250** 20 0.50 41 51

* Average test value for all samples

** Minimum test value for all samples

These results correlate with empirical correlations from similar rock types and other projects in the region within the same andesite rock formations. The fracture shear strength model was used in kinematic stability analyses of the bridge foundations. These analyses are discussed in detail in Section 4.

2.4.4 Rockmass Characterization

The Rock Mass Rating System (RMR) (AASHTO, 2010) and the Geologic Strength Index (GSI) (FHWA, 2006) systems were used to characterize the rock mass based on surface geologic mapping and boring information (window mapping, scan-line surveys, optical televiewer information, etc.). The RMR indicated values ranging from 59-72, with the average value of 65, indicating “good rock” by that classification system. The RMR results are presented in detail in Table 9.

Table 9:- Rock Mass Rating

TEST

HOLE

NUMBER

STRENGTH

OF INTACT

MATERIAL

DRILL

CORE

RQD

DISCONTINUITIES GROUND-

WATER TOTAL

SPACING CONDITION ORIENTATION

BH17-01 7 13 25 12 0 10 67

BH17-02 4 17 10 20 0 10 61

BH17-03 7 25 10 20 0 10 72

BH17-04 2 17 10 20 0 10 59

The GSI is a slight modification of the RMR system which is more appropriate for the bridge foundation evaluation at this site. The GSI is the algebraic sum of ratings assigned for the following rock mass properties:

• Intact rock strength

• Rock Quality Designation (RQD)

• Discontinuities spacing

• Discontinuities condition

• Presence of groundwater

For this project, GSI was estimated directly at rock outcrops by visual inspection and from the data logged from the boreholes. Specifically, the intact rock strength was estimated using the uniaxial compressive strength tests performed on core samples retrieved from the boreholes, the RQD, joint spacing, and condition of discontinuities were obtained from the borehole logs of the core. Groundwater was not encountered during the field investigations, however, as stated above, the GSI rating is determined with the groundwater rating set to 10 for all conditions. Table 10 provides a general description of the meaning of the GSI ratings.

Table 10: - General Meaning of GSI Rating Applied to Fractured Rock Masses

GSI RATING DESCRIPTION

COHESION OF

ROCK MASS

(psi)

FRICTION ANGLE OF

ROCK MASS

(degrees) 100 – 81 Very good rock >43.5 >45 80 – 61 Good rock 29 – 43.5 40 – 45 60 – 41 Fair rock 22 – 29 35 – 40 40 – 21 Poor rock 14.5 – 22 30 – 35

< 20 Very poor rock < 14.5 <30

From the estimated GSI values, the overall ranges of GSI for the north abutment and south abutment have been listed in Table 11. Estimation of rock mass in both the RMR and GSI systems had similar results, with an overall rating as “good rock”.

Table 11: - Estimated Range of GSI

LOCATION MEAN GSI HIGH GSI LOW GSI

Bridge #1520 64 67 61

Bridge #1521 65 72 59 All Data 65 72 59

Rock mass strength was determined using a modified form of the Hoek criterion discussed in Section 3.4 for intact rock strength (Hoek, 1988, 1994). For the rockmass, the criterion has the following form:

𝜎𝜎1′ = 𝜎𝜎3′ + 𝜎𝜎𝑐𝑐𝑐𝑐 �𝑚𝑚𝑏𝑏 ∗ 𝜎𝜎3′ 𝜎𝜎𝑐𝑐𝑐𝑐 + 𝑠𝑠� 𝛼𝛼 where 𝜎𝜎1′ = maximum effective principal stress 𝜎𝜎3′ = minimum effective principal stress 𝜎𝜎𝑐𝑐𝑐𝑐 𝑚𝑚𝑏𝑏 , 𝛼𝛼, 𝑠𝑠 uniaxial compressive strength of intact rock curve fitting coefficients related to rockmass quality

The rockmass coefficients, mb, and s, are determined from the following relations, which are functions of a rock mass quality classification parameter called the Geological Strength Index, GSI (Hoek 1994, Hoek and Brown 1997, Marinos and Hoek 2000). For disturbed rock masses, the following equation was used to determine the appropriate curve fitting coefficients:

mb = mi ∗ exp �

GSI − 100

for GSI > 25: s = exp �GSI−100 𝛼𝛼 = 0.5

Estimates of Mohr-Column rockmass failure envelopes are derived for various GSI value obtained for the site, the average or the minimum uniaxial compressive strength of the rock type at a foundation location, and the mi values derived for the rock fabric types and are presented in Table 12 below. These failure envelopes are directly related to bearing resistance discussed in the following Section.

Table 12:- Mohr-Coulomb Strength Parameters Calculated for the Rockmass Based on Hoek and Brown Criterion

GSI NUMBER σci (psi) s mb mi α

ESTIMATED ROCK MASS MOHR-

COULOMB

φ' c'

8,582*

0.0357 4.7930 20 0.500 54 269 70 0.0067 2.3464 20 0.500 52 138 60 0.0013 1.1487 20 0.500 48 82 50 0.0002 0.5623 20 0.500 43 55

3,250**

0.0357 4.793 20 0.500 50 128 70 0.0067 2.3464 20 0.500 46 75 60 0.0013 1.1487 20 0.500 41 51 50 0.0002 0.5623 20 0.500 35 37

* Average test value for all samples

** Minimum test value for all samples

SECTION 3--ANALYSIS AND RECOMMENDATIONS

This section provides a general interpretation of the site investigations and gives recommendations for design of the proposed structures. Generalized subsurface profiles were developed based on data collected during the investigation. A structural geology analysis and rockmass rating for the bedrock on which the bridge foundations will be constructed was analyzed and rockmass stabilization recommendations are provided. The structures considered for these sites are single spans bridges with concrete box girder.

Due to the competent nature of the bedrock and the low seismic hazard at the site, the bridges are anticipated to be supported on spread footings bearing on shallow bedrock. Although other retaining structures may be required, for this report recommendations for only semi-gravity cast-in-place walls will be included. Design for other wall types will be provided upon request.

3.1 SITE CHARACTERIZATION

The site investigation program suggests that the subsurface can be characterized for design purposes as a 2- to 5-foot thick layer of overburden granular soils (A-1-a and A-1-b) overlying andesite lava and flow-breccia bedrock to the maximum depth of exploration, approximately 28 feet. Due to the shallow elevation of the bedrock at the site and the existence of hard gravel and cobbles within the overburden soils, standard penetration tests (SPT) could not be performed, but the drilling pressure data suggests that the overburden material is medium dense to dense. A conservative corrected SPT-N160 value of 50 for the materials encountered in the overburden was assumed for abutment wall design. The encountered granular soil was estimated to have a conservative unit weight (γ) of 125 pounds per cubic foot, a friction angle (φ) of 33 degrees, and no cohesion. These values were used to estimate applied earth pressures.

Bedrock encountered in the borings is tan to dark brown and maroon, fine to medium grained, fresh to slightly weathered, and is strong and hard. The rockmass is fractured, mainly due to contraction that occurred during cooling of the lava flows. The fractures are widely to very closely spaced; range from low to high angle; slightly rough to rough; closed, healed, or open; with calcite infilling. The recovered core displayed a variable joint and fracture orientation in the rock mass with no foliation planes. Many of the joint sets within the rock appeared to be healed. Calcite and iron oxide deposits and staining were observed within the rock core samples indicating fracture flow groundwater conditions. Most of the rock core samples exhibited low to moderate fracturing with moderate to high RQD values.

Based on the results in Table 12, a combined minimum compressive strength value for the andesite and a GSI value of 60 results in a relatively conservative assumption for the rockmass properties at the bridge sites. Table 13 summarizes the rockmass physical properties that are recommended for the foundation design.

Table 133:- Estimated Rockmass Physical Parameters Used in Design

BEDROCK ROCK

TYPE

UCS

YOUNG’S

MODULUS

(psi)*

WET DENSITY

(pcf)

POISSON’S

RATIO

FRICTION

ANGLE

COHESION

Andesite 3,250 5.1x106 150 0.30 41 51

Groundwater was not observed in the test holes at the time of the subsurface investigation. Calcite mineralization and iron staining in the core samples indicates that fracture flow has likely occurred in the past. Fluctuations in groundwater levels due to seasonal and climatic effects are expected.

Groundwater is not expected to be encountered during installation of the shallow foundations elements for the proposed bridges.

3.2 DISCONTINUITIES AND KINEMATIC ANALYSIS

3.2.1 Proposed Abutment Foundations

The bridge abutments are planned to be supported on shallow foundations bearing on volcanic bedrock. Based on observations of surface outcrops near the proposed bridge locations, the principle fracture sets in the rock do not create geometries averse to construction of the bridge foundations. However, if during foundation excavation adverse bedrock joint sets became apparent, stabilization could be accomplished using shotcrete with steel mesh, and bolting rock blocks in place. If deemed necessary, adverse joint sets may be stabilized with rock anchors with unbounded lengths of about 15 feet. An ultimate ground-to-grout bond strength of 125 pounds per square inch should be used in stability calculations. A geotechnical engineer will be on site to inspect the foundation material stability and upon excavation completion and determine if additional spot bolting is required.

3.2.2 Reinforcement Design

For footings located on an inclined rock slope with unfavorable rockmass discontinuities, tensioned rock bolts may be required to secure footing stability. The bolting pattern and tension loads is generally design based on the fracture mapping and imposed structure loads. Transient uplift load resistance can be provided by the weight of the structure and tie-down anchors grouted into the underlying rock.

Based on limited observations in the field, as described in Section 2.1, no unfavorable rockmass discontinuities were identified. The geotechnical engineer should inspect the bridge abutment bearing excavation prior to construction to identify any unfavorable discontinuities present and provide direction on appropriate mitigation efforts. For the purposes of estimating, 20, #10 bars, 15-foot long rock bolts (with all accessories such as washers, nuts, plates) should be included in the contract. The exact number and bolt locations will be determined during construction following the foundation inspection upon excavation completion. A note should be incorporated in the plans indicating exact bolting locations and quantity will be determined during construction by a FHWA- CFL geotechnical engineer upon foundation excavation completion.

3.3 ABUTMENTS AND RETAINING WALL DESIGN RECOMMENDATIONS

A total of eight cast-in-place (CIP) semi-gravity walls are planned near the bridge abutments.

Based on the 70 percent plans, the proposed retaining walls varied in length between 13 and 38 feet with a maximum wall height of about 22 feet. Lists of the walls are shown in Table .

Table 14:- Retaining Wall Locations and Geometry

RETAINING

WALL NUMBER

APPROXIMATE

BEGIN STATION,

OFFSET

(ft)

APPROXIMATE

END STATION,

OFFSET

(ft)

LENGTH

(ft)

MAXIMUM

HEIGHT

(ft)

APPROXIMATE

BEARING

ELEVATION

(ft)

East Abutment Bridge #1520 Wingwall A

23+86, 17 LT 23+99, 17 LT 13 15 5,951.0

East Abutment Bridge #1520 Wingwall B

23+86, 17 RT 23+99, 17 RT 13 15 5,951.0

West Abutment Bridge #1520 Wingwall C

25+02, 17 LT 25+15, 17 LT 13 13 5,951.0

West Abutment Bridge #1520 Wingwall D

25+02, 17 RT 25+15, 17 RT 13 13 5,951.0

East Abutment Bridge #1521

NE Wing Wall

57+83, 17 LT 57+98, 17 LT 38 20 5,951.0

East Abutment Bridge #1521 SE Wing Wall

57+83, 17 RT 57+98, 17 RT 38 20 5,951.0

West Abutment Bridge #1521

NW Wing Wall

56+35, 17 LT 56+68, 17 LT 38 22 5,951.0

West Abutment Bridge #1521

SW Wing Wall

56+33, 17 RT 56+68, 17 RT 38 22 5,951.0

Overburden granular soils encountered near the proposed retaining wall locations generally consisted of non-plastic silty sand with gravel, cobbles, and boulders (A-1-a and A-1-b). Overburden material properties were estimated for wall design based on observations and geology, field investigation, and known empirical strength characteristics of similar materials. The modulus of elasticity for the soil (Es) is assumed to increase linearly with depth (z) at a rate of 560 pounds per square inch per foot of depth.

Table 5 provides a summary of the estimated overburden soil properties for design purposes.

Table 15:- Estimated Soil Properties for Design

SOIL MATRIX MATERIAL

DESCRIPTION

UNIT WEIGHT

γ, (pcf)

COHESION c, (psf)

FRICTION

ANGLE φ, (deg)

RATE OF SOIL

MODULUS

INCREASE nh, (psi/ft)

Silty SAND with Gravel (SM -GP-

GM)

AASHTO classifications A-1-a &A-1-b

125 0 33 560

3.3.1 Bearing Resistance for Abutments and Retaining Walls

All abutment footing locations are anticipated to be located on rock. At wall locations, it is anticipated that the walls will be founded on either rock or granular soil with sufficient bearing resistance. Based on the subsurface materials encountered at the project site and the methods used for determining soil strength parameters, the bearing resistance factors recommended for long-term stability condition were selected from the LRFD AASHTO Manual (2010), Table 10.5.5.2.2- 1 and are presented in Table 146.

Table 146:- Resistance Factors for Abutment and Wall Foundation Design

LIMIT STATE

RESISTANCE FACTORS, φ

BEARING RESISTANCE SHEAR RESISTANCE TO

SLIDING

PASSIVE PRESSURE

RESISTANCE TO

SLIDING

Strength 0.45 0.80 0.5

Service and Extreme Event 1.00 1.00 1.00

Strength I Limit State

For walls on soil, the bearing resistance at the Strength I Limit state was calculated for footings with a minimum 6-foot setback from existing slope to the toe of the footing using Meyerhof bearing capacity formula for continuous rigid footings in sand consistent with Article 10.6.3 of AASHTO Manual (2010). Zero embedment, no eccentricity, 45-degree front slope, and assumed material properties listed in

Table 5 were used to conservatively calculate the maximum factored bearing resistance of different footing widths on soil at the Strength I Limit state using a resistance factor of 0.45. The bearing capacity resistance factors were modified for footings on slopes. Table 157 presents the calculated factored bearing resistances for concrete footings bearing on soil with the AASHTO minimum recommended setback from existing slope to the toe of footing. Standard wall geometries shown in the table were provided by the FHWA-CFL structural engineer during the design process.

Table 157:- Bearing Resistance of Overburden Materials at the Strength Limit State

WALL HEIGHT

(ft)

FOOTING WIDTH

(ft)

BEARING RESISTANCE, AT THE

STRENGTH I LIMIT STATE

(ksf)

4.0 4.0 6.1

6.0 4.5 6.6

8.0 5.5 7.6

10.0 6.5 8.6

12.0 7.5 9.6

14.0 8.5 10.6

16.0 9.5 11.6

Calculated factored bearing resistance is anticipated to exceed the maximum factored bearing pressure imposed by the retaining walls. The width of the foundation for retaining walls should be a minimum of 2 feet. The footings should bear below the frost depth for local conditions which is estimated at 20 inches (below adjacent finish grade within 10 feet) for the project site. This is required to prevent heaving due to soil expansion from freezing and to prevent settlement due to loss of shear strength and stiffness from thawing. In sections where the walls bear on soil, the foundation should be prepared and compacted in accordance with Section 208 of FP-14 prior to constructing the footings. Soft or unsuitable soils are not anticipated, but if they are encountered, they should be over-excavated and replaced with compacted Foundation Fill in accordance with subsection 704.01 of FP-14.

At the bridge abutments bedrock is expected to be encountered beneath the overburden materials at relatively shallow depths (< 3 feet). Bedrock percent core recovery and RQDs were moderate to high for all completed core runs indicating “good” rockmass quality. Uniaxial compressive strengths of the rock samples at these wall locations obtained from laboratory tests are presented in Table 5. Footings bearing on bedrock that is stronger and less compressible than concrete are generally stable with negligible settlement. Due to the fractured nature of the rockmass, it is recommended that at the abutments and wall locations a nominal value of 225 pounds per square inch (32,400 pounds per square foot) with a resistance factor of 0.45 should be used when bearing on bedrock as the limiting bearing resistance at the Strength I limit state. The bearing resistance for the wall and foundation elements has been estimated based on the empirical rockmass strength criterion discussed in Section 2.4 and presented in Table 12 of this report. It was assumed that the at the required resistances within the rockmass to support the imposed loads at the strength and extreme event I limit states are defined by the "disturbed" parameters of the rock mass and the Hoek Criterion:

𝑞𝑞𝑢𝑢𝑢𝑢𝑡𝑡 = 𝜎𝜎𝑐𝑐𝑐𝑐�√𝑠𝑠 + �𝑚𝑚𝑏𝑏 √𝑠𝑠 + 𝑠𝑠� where, qult = Nominal Bearing Resistance at the abutments mb and s = Criterion parameters defined in Table 12 for disturbed rockmass σci = Uniaxial Compressive Strength of the intact Rockmass

A reduced setback length (4-foot setback) can be used for portions of the walls and the abutments where the footings are on bedrock. This is more economical than excavating to a typical wall footing width (6-foot setback) used for overburden soils. Table 16 provides an estimated foundation elevation for the design of abutment and wingwall footings based on the minimum recommended setback, foundation material types, and wall footing widths in accordance with the 70 percent plans.

Due to bedrock fracturing patterns, an irregular footing surface could occur along the retaining wall alignment in rock. Any depressions at the footing bearing surface resulting from removal of blocks should be leveled with lean concrete meeting Section 614 of FP-14. Grouting of fractures exposed during excavation of footings (dental grouting) should be performed with grout meeting the requirements of Section 725.13(b) of FP-14 for fractures less than approximately 1-inch wide;

and lean concrete meeting the requirements of Section 614 for fractures greater than 1-inch wide.

Service Limit State

The bearing resistance provided by the soil is a function of the amount of settlement. The amount of settlement of the walls is expected to be less than 1-inch when placed on native soils and less than 0.25 inch when placed on decomposed bedrock or on properly compacted foundation fill/lean concrete backfill. Based on the subsurface investigation, in certain locations, spread footing may be founded partially in rock and partially in soil with cobbles and boulders. In areas where the wall will bear on soil and bedrock along the entire footing, it is recommended that a minimum of 1-foot of the rock is sub-excavated below the bottom of the footing and laterally to a minimum of 2 feet beyond the foundation footprint and replaced with foundation fill compacted to a minimum of 95 percent relative compaction (ASTM D1557) to avoid adverse differential settlement. Footing foundations should be prepared in accordance with Section 208 of FP-14.

Lean concrete may also be used as backfilling beneath the footing elevation, however, the concrete surface shall be roughened prior to set-up by scarifying with a steel trowel be provide friction surface.

Extreme Event Limit State

Impacts of seismic loading (Kh) on wall stability are limited to lateral displacement and are not likely to cause wall collapse. Up to 2 inches of lateral displacement should be assumed during design considerations at the extreme event I limit state. The foundations are expected to perform satisfactorily when subject to a seismic loading (Kv) because a large margin of safety is provided for the bearing resistances at the static strength I limit state.

Lateral Earth Pressure

Bridge abutments and semi-gravity cantilever walls should be designed to resist lateral earth pressures and other applicable lateral loads in accordance with the AASHTO LRFD Specifications.

Lateral earth pressure acting on the back of a retaining wall is dependent on the following factors:

• Backfill material strength properties, including soil type and gradation, unit weight, moisture content, and friction angle.

• Backfill drainage.

• Soil-structure interaction.

• Compaction induced pressure and other surcharges.

• Earthquake loads

Backfill material strength properties are estimated based experience and on previous project practices and are provided in Table 16.

Table 16:- Recommended Wall Backfill Parameters for Design

MATERIAL

DESCRIPTION

UNIT WEIGHT γ, (pcf)

COHESION c, (psf)

FRICTION ANGLE φ, (deg)

Structural Backfill 150 750 35

Unclassified Borrow 125 125 33

Unbalanced groundwater behind walls and abutments adds significant lateral pressure and should be avoided by using permeable backfill and assuring a free draining gravity outlet for water.

Backfill in the immediate vicinity of the retaining structure should consist of structural backfill in accordance with subsection 704.04 of FP-14. Unclassified borrow may be used in the retaining zone behind the wall footing.

Rigid, absolutely retained walls that can tolerate little or no movement should be designed for the at-rest earth pressure (Ko) by an equivalent-fluid soil unit weight of 57 pounds per cubic foot for unclassified backfill material. That value is based on a compacted moist unit weight of 125 pounds per cubic foot and an effective friction angle of 33 degrees for leveled, free draining structural backfill, resulting in an at-rest earth pressure coefficient of 0.45

Retaining walls that can tolerate sufficient rotation at the top of the wall or lateral translation equal to or greater than 0.002 times the wall height should be designed for an active-lateral-earth pressure (Ka). The Rankine's lateral-active-earth pressure represented by an equivalent-fluid pressure of 37 pounds per cubic foot for leveled, free draining backfill can be assumed as the lateral earth pressure acting at a distance of one-third the wall height above the wall base.

The Rankine's method can be used if the “soil-structure interaction” is neglected. To account for the soil-structure interaction with a friction angle valued at 2/3 times the backfill effective friction angle (0.67 φ´), the Coulomb's method may be used for calculating the lateral-active- earth pressure.

In addition to the lateral earth pressure, retaining structures should be designed considering additional pressures due to traffic live-load surcharge equivalent to 2 feet of soil (250 pounds per cubic foot) as described in Article 3.11.6.4 of AASHTO LRFD (2010). To estimate the psudo-static coefficient (Kae) on retaining walls due to earthquake effects, the Mononobe-Okabe (M-O) pseudo-static theory should be utilized. The theory is applicable to walls which are free to yield sufficiently to permit development of lateral-active-earth pressure. For fixed-rigid walls that cannot rotate, the M-O theory will significantly underestimate the seismic force. A horizontal acceleration Kh equivalent to a site class “C” peak ground acceleration of 0.5As as described in Section 1.7 should be used in the pseudo-static analysis. Vertical acceleration coefficient (Kv) can be neglected when calculating lateral loads. A resistance factor of 1.00 should be assumed for the extreme event limit state.

When passive lateral earth pressures are applicable, wall movement of 0.02 times the wall height at the top of the wall are necessary to reach passive earth pressure. Procedures outlined in Appendix A11 of the AASHTO LRFD should be used during design to determine forces exerted on the walls.

Lateral loads imposed on the abutments and fill walls due to lateral earth pressure and horizontal seismic forces may be resisted by the friction developed between the cast-in-place concrete footing and the supporting soils with a resistance factor of 0.8. The nominal sliding resistance for concrete footings cast on cohesionless soil (native soil, foundation fill and decomposed/weathered bedrock) is 0.65. This factor is conservative and assumes cast-in-place concrete on sand. The passive pressure for free draining backfill can also be calculated using the aforementioned properties. For the passive pressure a resistance factor of 0.50 for LRFD is recommended. The wing wall design should consider the surcharge loading imposed from both traffic and construction equipment.

3.3.2 Earthwork Recommendations

It is anticipated that the planned fills will be constructed using locally available materials derived from the project excavations. The contractor may utilize portable screening plant and a crusher to remove fines and reduce over-sized rock within excavated materials to meet the select granular backfill specifications. If treated materials run out, importing of additional materials will be required. Based on recent experiences with similar materials, a shrink value of 10 percent is recommended as an earthwork factor to account for the volume change from excavation cut to the compacted embankment for the on-site granular soils. If andesite rock is processed for reuse in embankment construction, a bulking factor of 20 percent is recommended to account for volume change from excavation cut to compacted embankment. The construction techniques used in the field will have significant influence on the actual shrink/swell of the excavated materials. The in-place embankment densities will also be influenced by the gradation and screening of the fill materials, staging and placement, and method of compaction. In general excavation in competent bedrock should be minimized except for wall footing preparations.

The appropriate Occupational Safety and Health Administration and local regulations for health and safety for temporary excavation stability must be followed during construction. The overburden soil and fractured bedrock types should be classified by the contractor based on the boring logs and materials excavated on site. Any zone of intact stable rock should be evaluated by a qualified and a competent person before excavating slopes steeper than 1V:1H.

3.3.3 Erosion and Corrosion Protection

Riprap protection with erosion control mats are recommended in areas below wall drains and downspouts. Additionally, aggressive erosion protection and revegetation measures should be considered to assure stability of fill slopes in front of the walls.

Based on guidelines outlined in subsection 704.8(b) of FP-14, a pH below 5 or above 10, a resistivity below 3,000 ohm-cm and, a sulfate and chloride concentrations above 200 and 100 parts per million respectively, are representative of an aggressive soil environment. Due to the shallow bedrock encountered during the subsurface investigation, an adequate sample size of the overburden granular material could not be obtained for corrosion potential testing. However, the granular on-site soils are anticipated to have a low corrosion potential. Concrete mix proportions should be reviewed and approved by the FHWA-CFL Materials Group. No corrosion protection on the reinforcement steel in the walls is required if the minimum concrete cover of 3 inches is applied to protect the steel.

3.4 CUT SLOPE RECOMMENDATIONS

In general, the maximum recommended slope ratios for proposed cut slopes is 1V:1.5H for the cut slopes constructed in soil. For cut slopes constructed in rock, the maximum recommended slope ration is 2V:1H. The recommended slope ratio is dependent on the nature of the observed material and the existing stability of the slope. During the excavation of predominantly soil cut slopes on the project large boulders up to 10 feet (or larger) should be expected within the proposed excavation limits. Care should be taken to avoid over excavation by removing (or destabilizing) large boulders found to be crossing the proposed finished slope “neat line”.

A station-by-station recommended cut slope ratio is presented in

Table 17. These recommendations are based off of the 30 percent design alignment. Seepage was not noted during field exploration as described previously. If seepage is encountered during construction, the cut slope angle should be reduced to maintain a stable slope. For all soil slopes less than 5 feet high, a cut slope ratio of 1:2 (V:H) should be used.

Table 17:- Recommended Cut Slope Ratios by Station

Station to Station Side Slope Type Maximum Recommended Cut Slope Ratio (V:H)

52+00 - 56+50 LT

Volcanics, Flow Breccia and Andesite, tan to brown to maroon, fresh, medium strong to very strong. Joints close to moderately spaced, low to high angle, healed to closed, slightly rough to rough, calcite infilling.

2:1

Rock slopes should be provided with sufficient catchment ditches to reduce the risk of rockfall entering the traveled way.