UT_FLAP_149_(1)_FINAL_GEOTECH_REPORT.pdf

UT FLAP 149 (1)

BRUSH CREEK BRIDGE

REPLACEMENT PROJECT

Dinosaur National Monument

Access Road SR-149 Quarry Entrance Road, Uintah, UT

Final Geotechnical Report

Report # UT-SX-0149-15-01

Prepared by Federal Highway Administration

Central Federal Lands Highway Division

Geotechnical Services Branch February 2016

UT FLAP 149 (1) Brush Creek Bridge Page i

Table of Contents

SECTION ONE--INTRODUCTION

1.1 OBJECTIVE AND SCOPE

1.2 EXISTING BRIDGE CONDITIONS

1.3 PROPOSED BRIDGE FOUNDATION CONSTRUCTION

SECTION TWO--GEOLOGY AND SEISMICITY

2.1 GEOLOGY

2.2 GEOLOGIC HAZARDS

2.3 SEISMICITY

2.4 SEISMIC DESIGN PARAMETERS

SECTION THREE--SUBSURFACE INVESTIGATION

3.1 FIELD EXPLORATION

3.2 LABORATORY TESTING

SECTION FOUR--ANALYSIS AND RECOMMENDATIONS

4.1 BRIDGE FOUNDATION DESIGN

4.1.1 Geotechnical Profile

4.1.2 Groundwater Conditions

4.1.3 Brush Creek Bridge Foundation Selection

4.1.4 LRFD Driven Pile Foundation Design

4.1.5 Nominal Axial Resistance

4.1.6 Pile Setup

4.1.7 Group Effects on Axial Resistance

4.1.8 Lateral Loads

4.1.9 Pile Drivability

4.1.10 Settlement

4.1.11 Field Testing

4.2 ABUTMENT DESIGN

4.3 CORROSION POTENTIAL

4.4 CUT AND FILL SLOPE RECOMMENDATIONS

4.5 SHRINK/SWELL RECOMMENDATIONS

4.6 CONSTRUCTION CONSIDERATIONS

4.7 SPECIFICATIONS

4.8 LIMITATIONS

SECTION FIVE -- REFERENCES

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TABLES

TABLE 1 - Summary of Seismic Parameters Corrected for Site Class C TABLE 2 - Summary of Design Response Spectrum Data for Site Class C TABLE 3 - Summary of Brush Creek Bridge Borings TABLE 4 - Summary of Soil Classification Index Tests TABLE 5 - Summary of Electrochemical Properties TABLE 6 - Proposed Bridge Foundation Locations TABLE 7 - Estimated Subsurface Material Properties for Bridge Foundation Design TABLE 8 – N1(60) Calculation for Overburden Material at Bridge Location TABLE 9 - Summary of Design Loads for Brush Creek Bridge TABLE 10 - Nominal Axial Resistance (Required Resistance) for Pile Design TABLE 11 - Summary of Pile Recommendations TABLE 12 – LPILE (Lateral Load) Parameters TABLE 13 - Lateral Earth Pressures for Bridge Abutments TABLE 14 – Cut and Fill Slope Construction Recommendations

ILLUSTRATIONS

ILLUSTRATION 1 - Summary of Seismic Parameters Corrected for Site Class C

FIGURES

FIGURE 1 – Regional Location Map FIGURE 2 - Local Area Map FIGURE 3 - Geologic Map FIGURE 4 –Boring Logs Sheet

APPENDICES

APPENDIX A - Subsurface Investigation APPENDIX B - Laboratory Test Results APPENDIX C - Drilling Photographs and Notes APPENDIX D- A-Pile Analysis APPENDIX E – Pile Driving Analysis

UT FLAP 149 (1) Brush Creek Bridge Page 1

SECTION ONE--INTRODUCTION

This report presents the results of the geotechnical engineering investigation conducted for the Brush Creek Bridge, UT FLAP 149 (1) project. The Federal Highway Administration (FHWA) Central Federal Lands Highway Division (CFLHD), in cooperation with the National Park Service (NPS) and the Utah Department of Transportation (UDOT), is proposing to rehabilitate and widen a portion of the Utah State Route 149 [1]1, including a replacement bridge over Brush Creek located on this route, known as Quarry Entrance Road.

Utah State Route 149 (SR-149) is located in Uintah County, Utah approximately 10.4 miles southeast of Vernal, Utah as shown on “Regional Location Map”, “Local Area Map”, and “Geologic Map” on Figures 1 through 3, respectively. SR-149 is the main access into the western entrance of Dinosaur National Monument. The replacement bridge will be located on Quarry Entrance Road, immediately after intersecting with Brush Creek Road, approximately 100 feet southwest of the T junction.

The existing paved roadway has narrow shoulders ranging from zero to 2-foot narrow shoulders and inconsistent lane widths ranging from 10 to 13 feet wide. The intent of the project is to meet driver expectation by introducing a constant typical section and accommodate bicyclists by widening the shoulders. The project consists of the construction of a new two-lane bridge between Sta. 151+45.69 and Sta. 152+10.69, 6.5 feet left of the existing alignment, widening of approximately 4.1 miles of SR-149 to two 12-foot lanes with two 4-foot shoulders, drainage improvements, and grading.

1.1 OBJECTIVE AND SCOPE

The objectives of this investigation are to develop recommendations concerning structure foundations, subsurface drainage, corrosion potential, and shrink/swell values. In accomplishing these objectives, CFLHD conducted field exploration, laboratory testing, and engineering analysis.

1.2 EXISTING BRIDGE CONDITIONS

Based on information provided in the as-built drawings (Utah State Road Commission, 1954), the existing Brush Creek Bridge was constructed in 1954 as a single-span reinforced concrete slab on reinforced concrete full-height vertical abutment walls. The existing bridge is 35-foot long (clear span length) and 34-foot wide measured between outside edges of the deck. The clear roadway width between curbs is 28 feet. The substructure is founded on seventy 20-foot long HP10x42 steel H-piles.

1 Numbers in brackets refer to list of references.

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A scoping trip to the project site took place in July 18, 2013. The findings based on the bridge inspection [2] are summarized below:

The abutments and wing walls are in fair to good conditions. The wing walls at each corner of the bridge have rotated outward to a maximum distance of 1.25 inches creating a gap between the abutment and wing walls.

There are indications of bridge scour due to channel encroachment onto the southwest abutment. Stream flows have removed the stream bed material along the abutment exposing the pile cap at this location.

1.3 PROPOSED BRIDGE FOUNDATION CONSTRUCTION

The existing structure will be replaced by a two-lane bridge that meets the 7th Edition of AASHTO LRFD Bridge Design Specifications 2014 [3] and UDOT design criteria.

U-shape wing walls and pile bent (spill through) abutments with H-piles for the substructure are recommended. The new bridge will require the foundation to extend below the scour line. The pile bent (spill through) type abutments with driven piles are capable of addressing the scour demands.

Two-stage construction approach is recommended due to Right Of Way (ROW) limitations and the need to minimize site impacts while maintaining access. Staged construction will require temporary shoring and partial removal of the existing deck, approach slabs, abutments, and wing walls on one side and building enough width of the proposed bridge to accommodate a temporary lane configuration including a single travel lane and any space required for a temporary bridge barrier.

The first stage involves demolishing one half of the existing structure while keeping the traffic on a single lane on the other half. The new bridge and approach roadway is then constructed adjacent to the other half of the existing roadway and structure. The second stage will include shifting the single lane of traffic to the newly constructed section, demolishing the remaining structure, and constructing the other half.

The layout of the proposed bridge and site topography based on the 70 percent design plans, along with boring locations completed for this study, are presented in Figure 4.

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SECTION TWO--GEOLOGY AND SEISMICITY

2.1 GEOLOGY

Regional Geology

Dinosaur National Monument is located along the Utah-Colorado border on the southeast flank of the Uinta Mountains. The monument encompasses the confluence of the Green and Yampa Rivers and is dominated by several broad east-west oriented anticlines and synclines that expose Precambrian to Cretaceous sedimentary rocks. Tertiary and Quaternary surficial deposits occur on high mesas and along river valleys, respectively [2].

The project area is on the southwest margin of Dinosaur National Monument near the transition of the Split Mountain Anticline (north) to the Jensen Syncline (south). Geology adjacent to the roadway consists primarily of the Cretaceous Mancos Shale and Quaternary sediment and stream deposits (Hansen and others, 1991; Rowley and others, 1979).

Local Geology

The Mancos Shale dips approximately 50 degrees south near the north end of the alignment then flattens southward. Midway along the alignment, the roadway crosses the trough of the Jensen Syncline and the rocks are essentially flat lying. Continuing south, the Mancos Shale dips gently towards the north. From top to bottom, the Mancos Shale includes the Upper Shale, Frontier Sandstone, and Mowry Members. The Upper Shale is characterized by greenish-gray, fissile shale with some thin beds of siltstone, sandstone, and bentonitic clay.

The Frontier Sandstone is a light-greenish-gray, fine-grained calcareous sandstone with shale partings. The oldest unit, the Mowry Member, is characterized by bluish-gray, fissile, siliceous shale and bentonite with abundant fish scales [2].

Surficial deposits are unconsolidated or poorly consolidated patchy sediments. These sediments were deposited by the processes of weathering, mass wasting, and erosion and deposition by running water.

The majority of the road alignment traverses across surficial deposits associated with the Green River and its tributary, Brush Creek. These units include alluvium, floodplain and channel deposits, pediment deposits, and terrace deposits. Alluvium includes small deposits within and at the base of intermittent tributary channels; it consists of poorly-sorted sand, silt, and gravel up to about 20 feet thick.

Floodplain and channel deposits (Qfp) are the predominant surficial deposits along the Green River and Brush Creek. The deposits are mostly gravel and sand with localized pockets of cobbles and boulders, and may be up to 100 feet thick.

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The majority of the north half of the roadway alignment traverses pediment deposits characterized by poorly-sorted silt, sand, and gravel approximately 30 feet thick. The south half of the alignment crosses river terrace deposits that are remnants of alluvial deposits on benches along the Green River. The deposits are largely silt, sand, and gravel up to about 30 feet thick. Locally, this deposit may be cemented.

2.2 GEOLOGIC HAZARDS

Geologic hazards exist from both the natural environment of the project site and from existing and proposed construction of the roadway. Geologic hazards that exist in the vicinity of the project area are flash floods, highly erodible soils, and slope instability at fill slope locations.

Flash floods are possible in the creek drainage areas and at the roadway surface during periods of high precipitation. The risk is particularly high in the early spring months during snow melt.

Highly erodible embankment soils leading up to the bridge are susceptible to surface water erosion, which means particular attention should be paid to directing surface water flow on this project and constructing riprap rundown protection to protect the embankment.

2.3 SEISMICITY

Earthquake damage can be direct or indirect. Direct damage is due to strong shaking and fault rupture that occur mainly along faults in close proximity to the site during an earthquake.

Indirect damage can occur from ground shaking at a distance away from the site during an earthquake. There are no known active faults within the vicinity of the project site and it is geologically located in a low seismic area [4], as described further in the following section.

Therefore, neither direct nor indirect earthquake damage is anticipated at the project site. The nearest faults are approximately 14 miles north of the project site and include structures associated with the Diamond Gulch fault system, a series of southeast trending normal faults along the northern flank of Diamond Mountain in the eastern Uintah Mountains (Black and Hecker. 1999).

2.4 SEISMIC DESIGN PARAMETERS

The American Association of State Highway and Transportation Officials (AASHTO) criteria for seismic design of ordinary bridges, allows for damage to occur when the structure is subjected to 1000-year strong ground motion, but does not allow for collapse of one or more spans and loss of life. Bridges may suffer damage that requires repair or partial replacement.

However, AASHTO requires bridges to withstand small, more frequent earthquakes without damage. For Extreme Event I Limit State Design an earthquake load and resistance factor of 1.00 should be used during the LRFD design of the bridge structure.

Recommended seismic response parameters for the Brush Creek Bridge Project design are based

UT FLAP 149 (1) Brush Creek Bridge Page 5 on the (AASHTO) LRFD Bridge Design Specifications, 7th Edition, 2014 [4], and represents horizontal peak ground acceleration (PGA) with 7 percent probability of exceedance in 75 years (approximately 1000-year return period). The 1000-year return period uniform hazard spectrum for the bridge site, located at 40.40833º N latitude and -109.34042º W longitudinal, was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short- and long- period ground acceleration (Ss and S1 respectively) and corrected for the soil profile at the bridge site.

Ground motions caused by earthquakes are influenced not only by the distance from the fault planes, but also by the geology and soils found at the site. Subsurface profiles similar to the ones at the project site, with soil layers overlaying shale, will have amplified ground motions and a resonant period governed by the layer thickness and shear wave velocity of the materials. The damage potential of strong ground motions with respect to a structure is typically affected by the period of the strong earthquake motion and the resonant period of both the soil and the structure.

Based on the subsurface profile at the bridge site, the average time-weighted shear wave velocity for the top 100 feet (VS100) of subsurface materials was estimated between 1,200 and 2,500 feet per second. Therefore, the site is classified as Class C according to the site class definitions specified in Table 3.10.3.1-1 of AASHTO [3].

A seismic hazard analysis to establish ground motions for seismic design was conducted. The recommended spectral acceleration coefficient values for probabilistic design with a return period of 1000 years were calculated using the program provided with the AASHTO LRFD Bridge Design Manual developed by the USGS (2008) entitled “Seismic Design Parameters”, version 2.10, and are summarized in Table 1.

TABLE 1 - Summary of Seismic Parameters Corrected for Site Class C

Horizontal Peak Ground Acceleration, (As) 0.118g Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDs) 0.231g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.092g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.20 Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.20 Site Factor at Long-Period Range of Acceleration Spectrum, (Fv) 1.70

The 5% damped design response spectrum for Site Class C is shown in Illustration 1, Design Response Spectrum Corrected for Site Class C, and the supporting data is presented in Table 2.

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ILLUSTRATION 1 - Design Response Spectrum Corrected for Site Class C

TABLE 2 - Summary of Design Response Spectrum Data for Site Class C

Period (sec) Sa (g)

Sd (in)

0.00 0.118 0.000 (PGA)(Fpga) = As

0.10 0.231 0.014

0.20 0.231 0.090 (Ss)(Fa) = SDs

0.50 0.231 0.359

0.50 0.230 0.360

0.60 0.153 0.540

0.80 0.115 0.720

1.00 0.092 0.899 (S1)(Fv) = SD1

1.20 0.077 1.079

1.40 0.066 1.259

1.60 0.058 1.439

1.80 0.051 1.619

2.00 0.046 1.799

2.20 0.042 1.979

2.40 0.038 2.159

2.60 0.035 2.338

Based on the long acceleration coefficient SD1 value of 0.092g calculated as Fv.S1, the bridge site is assigned to Seismic Zone 1 according to Table 3.10.6-1 in AASHTO. Seismic hazard zones reflect the variation in seismic risk across the country and are used to permit different requirements for design as depicted in Table 4.7.4.3.1-1 in AASHTO.

0.00

0.05

0.10

0.15

0.20

0.25

0.0 0.5 1.0 1.5 2.0 2.5

Sp e ct ra l A cc e le ra ti o n , S a (g

Period, T (sec)

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SECTION THREE--SUBSURFACE INVESTIGATION

3.1 FIELD EXPLORATION

A field exploration program consisting of two subsurface borings was conducted by CFLHD in February 24, 2015. Drilling operations consisted of hollow stem auger, mud rotary, and wireline core drilling techniques. Borings were logged by representatives of CFLHD, and soil and rock core samples were visually classified in the field and shipped to CFLHD Materials Laboratory for testing.

Details of the field exploration program and the boring logs are included in Appendix A. Boring locations are presented on Figure 4. A summary of the boring locations and depths are shown in Table 3.

Table 3 - Summary of Brush Creek Bridge Borings

Borehole Number

Location Station Offset Ground

Elevation (ft) Depth Below

Ground Surface (ft)

B-101 Abutment 1 151+48 7.5 ft RT 4752.8 38.2

B-102 Abutment 2 152+08 8.5 ft LT 4753.4 33.2

At Abutment 1, below the 1.5-foot thick paved layer to 17.0 feet below ground surface (from El.

4751.3 to 4735.8), silty to clayey sand with gravel to well-graded gravel with fines were encountered. Cobbles were observed at the top of this 15.5-foot thick layer. The material was somewhat loose from 9 to 12 feet (from El. 4743.8 to 4740.8). Water table was encountered at

11.0 feet (El. 4741.8). Standard Penetration Test (SPT) blow counts were 17 at 6.3 feet (El.

4746.5) and 20 at 14.4 feet (El. 4738.4), indicating a medium dense layer. Weathered shale was encountered below this layer at 17.0 feet (El.4735.8) until the end of boring at 38.2 feet (El.

4714.6). All five SPT blow counts taken in the shale layer at 5-foot intervals reached 50 (refusal) at 2- to 4-inch depths, indicating a hard layer.

At Abutment 2, below the 1.5-foot thick paved layer to 17.0 feet below ground surface (from El.

4751.9 and 4736.4), varying amounts of gravel, sand, and fines were encountered. Gravel, sand, and fines with cobbles at top; sandy lean clay with gravel at middle; and silty to clayey sand with gravel at bottom were observed at this 15.5-foot thick layer. Water table was encountered at 10.0 feet (El. 4743.4). SPT blow counts were 8 at 9.5 feet (El. 4743.9) and 18 at 14.5 feet (El.

4738.9), indicating a loose to medium dense layer. Weathered shale was encountered below this layer at 17.0 feet (El.4736.4) until end of boring at 33.2 feet (El. 4720.2). SPT blow count taken at 19.5 feet (El. 4733.9) was 35, indicating a hard layer. The following three SPT blow counts taken in this layer at 4- to 5-foot intervals reached 50 (refusal) at 2- to 4-inch depths, also indicating a hard layer.

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3.2 LABORATORY TESTING

Laboratory testing was conducted on select soil samples obtained during the subsurface investigation to determine material properties for foundation design. Samples of similar material properties were combined for more adequate testing size. Laboratory test procedures and results are presented in Appendix B and summarized in Tables 4 and 5.

The overburden materials were classified as silty to clayey sand with gravel to well-graded gravel with fines (GW-GM) between 1.5 and 17 feet from the ground surface. Cobbles were encountered at top of this layer. Weathered shale was encountered in both borings at approximately 17 feet to end of borings. The shale samples were classified as lean clay (CL).

Table 4 - Summary of Soil Classification Index Test

Boring Number

Station

SPT

Sample Sample Depth Interval (feet)

Percent Passing

#200 Sieve

Water Content

LL PI USCS AASHTO

B-101 151+48

1 5.0-6.5 20 6.2 -- -- -- --

2 13.0-14.5 6.7 10.3 NV NP GW-GM A-1-a (0)

3 to 7 18.0-38.2 82 16.6 42 26 CL A-7-6 (21)

B-102 152+08

1 8.0-9.5 61 14.5 37 23 CL A-6 (11)

2 13.0-14.5 16 14.7 -- -- -- --

3 to 5 18.0-33.2 82 20.9 44 29 CL A-7-6 (23)

Notes: NV = No Value, NP = Non-Plastic, -- = Not Applicable/No test conducted

Electrochemical tests were conducted on a representative sample combined from both B-101 and B-102 to determine soil corrosivity on concrete and buried metals. Testing for resistivity and pH were performed in general accordance with AASHTO T 288 and T 289, respectively.

Due to low resistivity result (660 ohm-cm), sulfate and chloride contents were also tested in accordance with AASHTO T 290 and T 291, respectively. From the sulfate test, it may be concluded that concrete may suffer sulfate attack at a concentration of 0.317% sulfate content in the ground water. The end result of sulfate attack can be cracking and loss of strength. A summary of analytical test results are provided in Table 5.

Table 5 - Summary of Electrochemical Properties

Boring Number

Sample Depth Interval (feet)

Classification

(AASHTO)

(USCS)

pH

Resistivity (ohm-cm)

Sulfates (%/ppm)

Chlorides (%/ppm)

B-101 & B-102 Varies NA 7.6 660 0.317 / 3170 0.0144 / 144 Notes: NA = Not Applicable / No test conducted.

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SECTION FOUR--ANALYSIS AND RECOMMENDATIONS

For the analysis and foundation design of the proposed bridge, a geotechnical profile was developed from subsurface observations and laboratory testing conducted during the geotechnical investigation. Bridge foundation design, corrosion potential, shrink/swell recommendations, and construction considerations are discussed in the following sections. All foundations are in accordance with the American Association of State Highway and Transportation Officials (AASHTO) Probabilistic Load and Resistance Factor Design (LRFD), 7th Edition, 2014 [3].

4.1 BRIDGE FOUNDATION DESIGN

The proposed new structure is approximately 65-foot long and 40-foot wide simple-span bridge with prestressed box beam girders and cast-in-place concrete deck [5]. The new bridge will be located 6.5 feet left of the existing alignment and at the same approximate elevation near the existing embankment fills at both abutment locations. The elevation of the proposed bridge deck at centerline is 4752.8 at Abutment 1 and 4753.4 at Abutment 2, with an average elevation of

4753.1 at the center of the bridge. The channel invert elevation is 4741.6 at the centerline of the proposed bridge. The proposed abutment centerline stations and pile cap elevations are shown in Table 6.

TABLE 6 - Proposed Bridge Foundation Locations

Foundation Approximate

Station, ft Top of Pile Cap

Elevation Bottom of Pile Cap

Elevation Top of Pile Elevation

Abutment 1 Sta. 151+45.69 4751.29 4747.29 4749.29

Abutment 2 Sta. 152+10.69 4752.80 4748.08 4750.08

Note: Top and bottom pile cap elevations are based on 95 percent plan drawings dated February 2016.

4.1.1 Geotechnical Profile

Material properties were estimated for foundation design based on the field investigation results, laboratory test results, and presumptive empirical strength characteristics of similar materials due to similarities in the two abutment borings, the more conservative profile was selected for the design profile. Selected design values are presented in Table 7.

In general, two distinctive soil layers, overburden material and weathered shale (cohesive IGM), were encountered in each of the abutment borings at similar elevations. These materials were generally characterized as gravel, sand, and fines with cobbles as overburden material at top layer and weathered shale at bottom layer.

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TABLE 7 - Estimated Subsurface Material Parameters for Bridge Foundation Design

Depth Below

Ground Surface

(ft)

Material Description

Unit Weight ɣ (pcf)

Effective Unit

Weight1 ɣ’ (pcf)

Undrained Shear

Strength (Su)2 c (psf)

Friction Angle

Փ (deg)

Rate of Increase of Soil Modulus with Depth for Sand3 nh (ksi/ft)

Soil Modulus for Sand Ks (pci)

1.5-10.0 Gravel, sand, and fines with cobbles at top

120 N/A 0 32

1.11 for dry or moist, medium sand 92.5

10.0-17.0 Gravel, sand, and fines 120 57.6 0 32

0.556 for submerged, medium sand 46.3

17.0-38.0

Moderately to Slightly

Weathered Shale

120 120 10,0004 05 N/A N/A

1Groundwater is estimated at 11 feet (Abutment 1) and 10 feet (Abutment 2). Effective unit weight is used below the groundwater surface.

2The undrained shear strength (Su) can be estimated for low plasticity clays (PI≤10) and medium to high plasticity clays (11≤PI≤40) using the relationship developed by McGregor and Duncan, 1986.

3The modulus of elasticity for soil (Es) can be assumed to increase linearly with depth starting at zero at the ground surface. The rate of increase of soil modulus (nh) is applied per foot of overburden (z) as shown in the equation Es=nh*z (AASHTO 2014, 7th Edition).

4Su=0.15*N60 for medium to high plasticity clays. N60 >50 blows/foot (McGregor and Duncan, 1986) 5Typcially, the total internal friction angle () is negligible and assumed to be zero (=0) in cohesive materials.

Moderately to slightly weathered shale consistently measuring greater than 50 blows per foot were encountered at 18.8 to 19.5 feet to end of borings. The shale samples were classified CL in accordance with the Unified Soil Classification System and as A-7-6 in accordance with the AASHTO soil classification system.

The SPT-N values were measured every 5 feet and corrected for a hammer efficiency of 80 percent (assumed) and overburden pressure to obtain the N1(60) values for LRFD foundation design. A summary of the steps used for determining a representative N1(60) value for deep foundation design purposes are shown in Table 8. The groundwater table was encountered at an approximate depth of 10 to 11 feet from the ground surface at the time of drilling, therefore, pore-water pressure is partially included in calculating the effective overburden stress below 10 feet to 17 feet depth from the ground surface. Water table was not taken into account in calculation of the overburden stress of shale as shown in Table 8.

SPT N1(60) values within the overburden materials range from 12 to 30 with the higher values occurring when gravels were encountered. In general, SPT N-values obtained within shale reached 50 blows prior to penetrating 6 inches (indicative of very stiff material). In cohesive soils, SPTs are not very reliable for predicting strength and compressibility. Fine-grained, cohesive materials show different driving resistances when dry or moist, and if they become highly moist, they may not show the stiffness predicted by the SPT. From the field observations

UT FLAP 149 (1) Brush Creek Bridge Page 11 during subsurface explorations, the bedrock is expected to be a suitable foundation material for the H-piles.

TABLE 8 – N1(60) Calculation for Overburden Material at Bridge Location

Boring SPT Depth1, ft σv' 2,ksf N3 N60

Cn

5 N1(60)

N1(60)

Less Refusal

B-101 (Top El.

4752.8)

1 6 720 17 22.7 1.343 30.5 30

2 14 1,493 20 26.7 1.100 29.3 29

3 18 2,160 Shale Shale 0.976 Shale Shale

4 23 2,760 Shale Shale 0.894 Shale Shale

5 28 3,360 Shale Shale 0.828 Shale Shale

6 33 3,960 Shale Shale 0.773 Shale Shale

7 38 4,560 Shale Shale 0.726 Shale Shale

B-102 (Top El.

4753.4)

1 9 1,080 8 10.7 1.208 12.9 12

2 14 1,430 18 24.0 1.114 26.7 26

3 19 2,280 35 46.7 0.958 44.7 Shale

4 23 2,760 Shale Shale 0.894 Shale Shale

5 28 3,360 Shale Shale 0.828 Shale Shale

6 33 3,960 Shale Shale 0.773 Shale Shale

1Approximate mid-depth of SPT-N reading 2Effective overburden pressure 3Field SPT-N value 4Calculated using AASHTO equation 10.4.6.2.4-2 5Calculated using AASHTO equation 10.4.6.2.4-1

6Calculated using AASHTO equation 10.4.6.2.4-3 7Calculated using AASHTO equation 10.4.6.2.4-4.

Refusal values neglected in determining average N1(60)

The SPT reached 50 blows prior to penetrating 6 inches – these values were neglected.

4.1.2 Groundwater Conditions

Groundwater elevations are generally required for evaluating side and tip resistance under axial loads, for computing lateral loads, and for determining driven pile construction procedures in cohesionless soils. Groundwater was encountered during drilling close to the elevation of the stream channel, as indicated on the boring logs in Appendix A. Fluctuations in the groundwater level due to seasonal and climatic effects was not monitored, but should be expected. Based on the subsurface investigation and site observations, groundwater should be anticipated during foundation construction and pile driving operations. For design purposes, groundwater was assumed to be at 10 feet below the ground surface.

4.1.3 Brush Creek Bridge Foundation Selection

Driven (non-displacement) steel H-piles were selected for the foundation of this bridge after other options were considered. A shallow foundation system was determined unfeasible due to the depth from the abutment elevation to the stream channel and potential depth of design scour.

Estimated scour depths provided by CFL Hydraulic Engineer include the design scour (100-year)

UT FLAP 149 (1) Brush Creek Bridge Page 12 elevation of 4724.7 and the check scour (200-year) elevation of 4720.1; however, based on practical refusal (blows per foot larger than 50) in shale, non-scourable depth of approximately elevation 4734 was selected for the bridge design calculations for both geotechnical pile resistance and lateral pile resistance.

Deep foundation options were evaluated based on feasibility and costs. Driven piles, drilled shafts, and micropiles were all considered feasible at this project site based on site subsurface materials consisting of gravel, sand, and fines with up to 1-foot size cobbles in the stream channel. Based on local construction practices and foundation costs, driven steel H-piles are recommended for the foundation at the bridge abutments. Despite the possible encounter of cobbles in the overburden material, it is anticipated that pile can be driven to design depth within the shale.

Driven piles are capable of supporting heavy concentrated loads because they resist axial and lateral loads. Piles are designed to have adequate geotechnical and structural axial and lateral resistances for a tolerable settlement provided by CFL bridge engineers [6].

4.1.4 LRFD Driven Pile Foundation Design

Driven steel HP12x84 H-piles will be the main foundation system for the bridge abutments. The piles will gain axial resistance from a combination of side and tip resistance within the shale below the scour depth.

Based on the AASHTO LRFD (Load and Resistance Factor Design) guidelines, nominal axial resistance and loads should be factored by the appropriate limit state resistance factor (φ) for strength, service, and extreme event, and be compared to factored resistances. Factored resistances are determined from estimated nominal pile resistances and resistance factors that are based on the construction methods and testing, and monitoring (static load test, dynamic analysis, and wave equation analysis).

Table 9 presents the service loads and factored loads for each of the abutments.

Table 9 - Summary of Design Loads for Brush Creek Bridge

Location Maximum Service Load per Pile (kips)

(Service I Limit State*)

Maximum Factored Load per Pile (kips)

(Strength I Limit State)

Abutment 1 165 kips 225 kips

Abutment 2 165 kips 225 kips

*Service I Limit State loads have load factors of 1.0 and, therefore, the same as the unfactored loads.

The proposed piles are designed to resist design load through side and tip resistances developed within the overburden materials and shale. It is assumed at least two piles (one at each abutment) will be installed using pile dynamic testing (stress wave measurements) with a Pile Driving

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Analyzer (PDA) and signal matching with the best estimates of nominal resistance made from a restrike. Therefore, in accordance with the AASHTO “LRFD Bridge Design Specifications” 7th Edition [4] Table 10.5.5.2.3-1, the resistance factor for design will be 0.65.

Based on the factored loads in Table 9 and the design resistance factor, the nominal axial resistance required to resist for the static pile resistance analysis is summarized in Table 10.

Table 10 – Nominal Axial Resistance (Required Resistance) for Pile Design

Location Nominal Bearing Resistance (Target Resistance) per Pile (kips)

Abutment 1 347 kips

Abutment 2 347 kips

If the pile driving analyzer (PDA) is not used during installation, the resistance factor used to determine the minimum nominal pile capacities will need to be modified.

4.1.5 Nominal Axial Resistance

The piles will need to extend through overburden soils and develop resistance in shale driven to the minimum tip elevation due to lateral load resistance calculation performed by CFL Bridge.

The geotechnical resistance should also be confirmed by dynamic testing, i.e., the nominal geotechnical resistance indicated by the dynamic testing multiplied by the appropriate resistance factor should be greater than or equal to the factored structural axial resistance of the pile section. Assuming dynamic testing is completed as described in Section 4.1.4, a resistance factor of 0.65 on the geotechnical resistance indicated by the dynamic testing is appropriate (AASHTO Table 10.5.5.2.3-1, 2014). A resistance factor of 1.0 is recommended for the service and extreme event I limit states.

Pile installations will likely encounter hard driving within the overburden material. The overburden material also contains cobbles and possibly boulders that, if encountered, may result in difficult and slow driving conditions during the installation. We recommend that a protective rock tip be welded to the end of each pile to reduce the possibility of pile damage during driving.

The proposed HP12x84 steel H-pile sections were analyzed with the resistance gained through both frictional side resistance and end bearing resistance using the geotechnical profile material properties presented in Table 7. The box perimeter and box area were used for pile perimeter and tip area calculations, respectively.

Due to potential channel degradation scour at the bridge site, the reduction in pile resistance was considered by removing all of the alluvial material on top of the shale. This assumption accounts for loss of pile resistance in the scour zone as well as the influence of the reduced effective overburden pressure.

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No geotechnical losses were used to calculate the driving resistance, and were assumed to be negligible. Additionally, no downdrag or uplift forces were considered in the pile design based on observations from drilling on-site, laboratory test results, and experience working with shale materials from the same geologic unit. Table 11 summarizes the pile recommendations based on the pile analysis, The foundation analysis was performed using the APILE software [7] based on the loading determined from CFL Bridge Engineer, and the scour depth interpreted from practical refusal (blows per foot larger than 50) in shale.

Table 11 - Summary of Pile Recommendations

Recommendations Abutments 1 and 2

Pile Resistance Verification

At least two dynamic testing with PDA and signal matching at end of driving one at each abutment

Resistance Factor for Driven Piles 0.65

Pile Type HP12x84 H-Pile

Pile Material Steel, Fy=50 ksi min.

Number of Piles per Abutment 7

Compressive Nominal Axial Resistance (Target Resistance) Required During Driving (Includes Tip Resistance plus Side Resistance)

347 kips

Bottom of Pile Cap Elevation

4747.29 (Abut. 1)

4748.08 (Abut. 2)

Geotechnical Minimum Pile Penetration Depth1

18 feet in shale

Estimated Length of Piling to Achieve Required Nominal Axial Resistance (including embedment of 2 feet into each pile cap)

37 feet

Estimated Pile Tip Elevation2 4718.1 (or deeper)

1Minimum depth below ground surface, scour elevation, or bottom of excavation, to which a pile must be driven 2Minimum elevation to which piles must be driven to assure geotechnical design requirements are met.

The final pile length will be established in the field using the recommended dynamic field verification method. From the geotechnical perspective, the minimum pile tip elevation is required to be 18 feet in shale. The lateral load analysis conducted by CFL Bridge Engineer confirmed the minimum tip elevation to be at 35 feet (El. 4718.1) to prevent significant horizontal movements of the pile group.

4.1.6 Pile Setup

Pile setup is an increase in the nominal axial resistance that develops over time, predominantly

UT FLAP 149 (1) Brush Creek Bridge Page 15 along the pile shaft. When soils are compressed and disturbed due to pile driving, large excess pore pressures develop. These excess pore pressures are generated partly from the shearing and remolding of the soil and partly from radial compression as the pile displaces the soil. The excess pore pressures cause a reduction in the effective stresses acting on the pile, and thus a reduction in the soil shear strength. This results in a reduced pile resistance during, and for a period of time after, driving. After driving, the excess pore pressures will dissipate primarily through radial flow of the pore water away from the pile. With the dissipation of pore pressures, the soil reconsolidates and increases in shear strength. This increase in soil shear strength results in an increase in the static pile resistance and is called soil setup.

It is recommended that the resistance gained by setup not be utilized for determining nominal resistance for the pile foundation at the Brush Creek Bridge due to the shale encountered.

4.1.7 Group Effects on Axial Resistance

The resistance of a pile group to the applied axial loads is not necessarily the sum of the axial resistance of individual piles within the group. The zone of influence from an individual pile in a pile group may overlap with other piles, depending on the pile spacing. Historically the efficiency of groups of piles has not been a concern as long as the center-to-center spacing between piles is greater than 2.5 times the pile diameter (2.5B) or 30.0 inches whichever is smaller to avoid interference between adjacent piles during construction. Therefore, an efficiency factor (η) of 1.0 is applied for the pile group with center-to-center spacing of 2.5 diameters or more. It is anticipated that there will be seven piles at each abutment spaced at 6 feet apart, which corresponds to an efficiency factor of 1.0. Embedment of the piles into the pile cap should be at least 12.0 inches to meet AASHTO requirements. The anticipated pile cap embedment is

2.0 feet determined by the Bridge Engineer.

4.1.8 Lateral Loads

Lateral soil-structure interaction analysis of the driven piles is performed by CFL Bridge Engineer on single and group piles. For lateral loading analyses, use the subsurface soil parameters provided in Table 7 to estimate the lateral load displacement behavior. The values presented in Table 7 do not include reduction factors or load factors. A strength limit state resistance factor of 1.0 is specified in AASHTO for lateral geotechnical resistance of a single pile or pile group. Factored strength limit state lateral loads should be used in the analysis. A lateral load analysis should be conducted on a single pile from each abutment using the soil parameters presented in Table 12.

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Table 12 – LPILE (Lateral Load) Parameters

Soil Type Weak Rock

Effective Unit Weight (’), pcf 120

Elastic Modulus (E), psi 0.25 x 106

Uniaxial Compressive Strength (UCS), psi 1,000

Average RQD 25

Softness Constant (Krm), pci 0.0005

The soil behavior around the laterally loaded pile is described by the soil P-y curves. Material properties in Table 7 are for single piles and do not account for the reduced lateral resistance of piles in a group. P-multipliers are a function of the number of rows of piles and center-to-center pile spacing in the direction of loading. P-multipliers are required even for a single row of piles if the center-to-center spacing of the piles is less than 5 pile diameters. P-multipliers are specified in Table 10-7.2.4-1 in AASHTO. When this analysis method is used, the resistances at the strength limit state as represented by the P-y curves should not be factored since they already represent the nominal conditions.

4.1.9 Pile Drivability

A pile must satisfy two aspects of drivability. First, the pile must have sufficient stiffness to transmit driving forces large enough to overcome soil resistance. Second, the pile must have sufficient structural strength to withstand the driving forces without damage. A pile with a higher pile material strength can tolerate higher driving stresses that may allow a larger pile hammer to be used. This may allow a slightly higher resistance to be obtained before refusal driving conditions or pile damage occurs. Even if the pile structural resistance and geotechnical resistance both indicate a high pile resistance could be used, a high pile resistance may still not be obtainable because driving stresses may exceed allowable driving stress limits. A pile cannot be driven to a nominal static resistance that is as high as the structural resistance of the pile because of the additional dynamic resistance or damping forces generated during pile driving.

Pile drivability analysis was conducted using the Wave Equation Analysis Program (WEAP) and is included in Appendix E. GRLWEAP (PDI, 2010) computer software analysis [8] was performed to determine the drivability of the recommended pile tip elevations without damage.

GRLWEAP is a one-dimensional WEAP that simulates the pile response to pile driving equipment. At each abutment location, the static capacities from the APILE analysis were used along with the following parameters:

HP12x84 steel H-pile driven approximately 35 feet in length;

Maximum allowable pile driving stress of 45 ksi (90% of yield strength of steel piles);

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Blow counts ranging between 3 to 5 blows per inch;

A Del-Mag D 30-32 hammer with rated energy of 75.4 ft-kips;

Stroke of 8.84 feet;

Hammer efficiency of 80 percent;

Hammer cushion and pile cushion consisting of 1-inch of aluminum and 1-inch of Conbest, for a total thickness of 2 inches with an area of 1,000 in2; and

Pile cap weight of 1.9 kips.

WEAP results indicate that the selected H-pile will be drivable, using a hammer of 75.4 ft-kips, to the required penetration elevation. The analyzed hammer will drive the pile with blow counts ranging from 3 to 5 blows per inch and without exceeding the maximum driving stress, 45 ksi.

Driving conditions were also modeled using APILE with no scour. An axial resistance versus pile depth plot was obtained from the pile analysis. The design nominal axial resistance of 347 kips corresponded to a depth of 31.0 feet which is less than the minimum tip elevation. Design depth of 35.0 feet corresponded to 413 kips ultimate resistance which is less than 45 ksi pile driving stress for the entire driving operation. The APILE with no scour results confirm the WEAP analysis, and are presented in Appendix E, page E-5.

Prior to driving the test piles, the contractor should perform a WEAP analysis for the proposed driving setup. The WEAP analysis and pile driving operations should be in accordance with FP- 14 Section 551- Driven Piles [9]. A pile driving analysis should be conducted during installation of the piles to verify these preliminary analyses and protect the integrity of the piles. The contractor should not exceed the allowable driving stresses during installation. There can be variations in the subsurface conditions from those encountered during the field investigation.

4.1.10 Settlement

Pile group settlements were analyzed using the equivalent footing method. Resistance factor of

1.0 for the service limit state is recommended to assess the ability of the foundation to meet the specified deflection criteria. Settlements are expected to be minimal, less than 1-inch.

4.1.11 Field Testing

As stated in Section 10.7.9 of the LRFD manual [3], it is recommended a probe (test) pile be driven at the location of each abutment to:

1. Determine pile installation characteristics;

2. Evaluate the pile resistance with depth; and,

3. Establish driving criteria for the remainder of the piles.

The first driven pile at each abutment will be considered the test pile. Due to the small number of piles, it will not be necessary to have an initial, sacrificial test pile. If the first test pile meets requirements, then it can be incorporated as a production pile, which will save the construction costs and time.

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For the pile design, it was assumed at least two piles (one at each abutment) will be installed using dynamic testing (stress wave measurements) with a PDA and signal matching with the best estimates of nominal resistance.

4.2 ABUTMENT DESIGN

Abutments and wing walls should be designed to resist lateral earth pressures and other applicable lateral loads in accordance with AASHTO. Lateral earth pressure is influenced by the strength of the abutment backfill, the presence or absence of water, and the ability of the abutment or wall to move in response to lateral loads. Other loads, such as live loads, construction loads, and soil compaction loads should also be considered in the design.

Unbalanced water behind an abutment or wall adds significant lateral pressure and should be avoided by using free draining gravity outlets for water. Abutment and wing wall backfill should consist of select granular backfill (SCR, Section 704.10).

The coefficient of at-rest earth pressure should be used for design if the abutment is so restrained that it cannot be expected to rotate (deflect at the top) 0.002 times the wall height. Where deflection of the abutment can be expected, a coefficient of active earth pressure should be used for wall design. Active and at-rest lateral earth pressures of native materials, properly placed and compacted select granular backfill, and unclassified borrow above the water table are presented in Table 13. The values are unfactored loads and assume that the surface of the soil slope behind the wall is horizontal.

Table 13 - Lateral Earth Pressures for Bridge Abutments

Backfill Type Assumed Backfill

Properties Case

Unfactored Equivalent Fluid Density (pcf)

Select Granular Backfill c = 0 psf = 34 deg.

γ = 125 pcf

Active 35

At-Rest 55

Unclassified Borrow c = 0 psf = 30 deg.

γ = 125 pcf

Active 42

At-Rest 63

Silty to Clayey Sand with Gravel (SC-SM) c = 0 psf = 32 deg γ = 120 pcf

Active 37

At-Rest 58

4.3 CORROSION POTENTIAL

Acceptable ranges of soil aggressiveness on anticipated steel and concrete structures vary with respect to the proposed structure. For concrete and buried steel structural and drainage elements, tested soil sample exhibited high corrosive rate. A summary of analytical test results are provided in Table 5.

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The laboratory electrical resistivity test results indicate a 660 ohm-centimeters soil resistivity value which is significantly less than 5,000 ohm-centimeters, i.e., severe corrosive ground conditions for steel piles. The structural impact of corrosive soils on steel piles is calculated based on the corrosion rate multiplied by the surface area of the pile. For the pile life, the ratio of surface area to weight has major impact on time it takes for corrosion in a pile to reach the design weight limit. The proposed pile shape to be used for the bridge foundation also affects corrosion. All surfaces of the H-pile will be exposed to the aggressive soils, and these piles generally corrode faster comparing to round piles. A corrosion control using thickness allowance for corrosion is recommended, with sacrificial steel thickness of 0.15-inch is recommended to account for potential corrosion loss. It is anticipated that the recommended H-pile section will account for this corrosion loss without a significant loss in geotechnical or structural resistance.

The concentration of water soluble sulfates represents a severe degree of sulfate attack on concrete exposed to these materials. The degree of attack is based on a range of negligible, moderate, severe, and very severe as presented in the sulfate table 4.3.1 from American Concrete Institute (ACI, 2010) Building Code Requirements for Structural Concrete [10]. Based on this information, special sulfate resistant cement may be required for concrete exposed to the on-site soils based on the recommendations of a corrosion specialist. Concrete structures in contact with onsite soils, including abutments and wing walls, should be evaluated by a corrosion specialist to determine the appropriate mix design for this project.

4.4 CUT AND FILL SLOPE RECOMMENDATIONS

The slope ratio recommendation is dependent on the nature of the observed material and the existing stability of the slope. Ideally, a slope ratio of 1V:2H or flatter is recommended for proposed fill slopes to promote vegetation growth and stabilize the fill slope surface. The proposed fill slope ratio is not applicable due to the Right Of Way (ROW) constraints at the specified location, therefore shoulder stabilization is recommended on the steep fill slopes for the stability of fill slopes less than 8-foot high when the design slope ratio is steeper than the recommended fill slope ratio of 1V:2H. A maximum slope ratio of 1V:1.5H is recommended for the proposed cut slopes constructed in clayey soil with gravel and cobbles. Based on observations on-site, rockery construction is recommended for cut slopes less than 8-foot high when the design slope ratio exceeds the recommended cut slope ratio of 1V:1.5H. The construction recommendations for the proposed cut and fill slopes are listed in Table 14 below.

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Table 14 –Cut and Fill Slope Recommendations

Station Existing Condition Maximum Slope Ratio

Construction Recommendations

150+00 to 152+75 RT Existing ROW does not allow for flatter fill slope construction

1V:2H Fill Shoulder Stabilization

Drwg 251-C

158+50 to 164+50 LT Steep existing cut slope 1V:1.5H Cut Rockery Drwg 252-A

166+50 to 169+00 LT Steep existing cut slope Approach road…