Hanapepe_Draft_Final_Geotechnical_Report_ver4.pdf

D ra f t Repor t (ve r .4)

Final Geotechnical Report

Hanapepe River Bridge Kaumualii Highway, Route 50

Kauai, Hawaii State of Hawaii

Prepared for Federal Highway Administration

Central Federal Lands Highway Division 12300 West Dakota Avenue

Lakewood, CO 80228

August 2016

Hirata & Associates, Inc.

99-1433 Koaha Place

Aiea, HI 96701

9191 South Jamaica Street

Englewood, CO 80112-5946

W.O. 15-5683.1 Page ii

Contents Section Page

1.0 Introduction

1.1 Background and Location

1.2 Purpose and Scope

2.0 Geology and Seismicity

2.1 Regional Geology

2.2 Seismic Design Parameters

3.0 General Subsurface Characterization

4.0 Analysis and Recommendations

4.1 Bridge Foundation

4.1.1 Drilled Shaft Axial Capacity

4.1.2 Drilled Shaft Lateral Load Analysis

4.1.3 Drilled Shaft Construction

4.1.4 Trial Shafts and Load Test

4.1.5 Integrity Testing

4.2 Abutment Walls and Wing Walls

4.3 Retaining Walls

4.3.1 Spread Footing Bearing Resistance

4.3.2 Footing Lateral Resistance

4.4 Lateral Design

4.4.1 Passive Earth Pressures

4.4.2 Active and At-rest Earth Pressures

4.4.3 Dynamic Lateral Earth Forces

4.4.4 Drainage

4.5 Bridge Approach Slab

4.6 Pavement Section

4.7 Site Grading

4.7.1 Site Preparation

4.7.2 Structural Excavations

4.7.3 Embankments

4.7.4 Slope Gradients

4.7.5 Onsite Fill Material

4.7.6 Imported Fill Material

4.7.7 Backfill for Bridge Abutment Walls and Wing Walls

4.7.8 Compaction

5.0 Special Contract Requirements

6.0 Additional Services

7.0 Limitations

W.O. 15-5683.1 Page iii

Appendixes

A Field Exploration Program & Boring Logs

B Laboratory Testing

C P-Y Curves Data

HIRATA & ASSOCIATES, INC.

W.O. 15-5683.1 Page 1

1.0 Introduction

1.1 Background and Location

This report provides geotechnical recommendations for the proposed Hanapepe River Bridge replacement located along Kaumualii Highway, Route 50, in the District of Waimea, Kauai, Hawaii. Hanapepe River Bridge is located at MP 16.57, approximately 150 feet west of the Iona Road and Kaumualii Highway intersection.

Project vicinity and bridge photos are presented on Figures 1 and 2.

FIGURE 1

Project Vicinity Map

The Central Federal Lands Highway Division (CFLHD) of the Federal Highway Administration (FHWA), in cooperation with the Hawaii Department of Transportation is proposing to replace the existing approximately 275-feet long, 35.8-feet wide, three span concrete bridge over Hanapepe River to meet current design standards for roadway width, load capacity, bridge railing and transitions, bridge approaches, and to mitigate the effects of scour.

The existing bridge was constructed in 1938, with the abutments and piers supported on timber piles.

Based on as-built plans, the timber piles were estimated to be about 30 feet in length. Undermining was reported at both intermediate piers and deterioration was observed on the exposed timber piles below the pile caps. However, no noticeable distresses due to settlement of the pile foundations were observed.

The replacement bridge will consist of a three span concrete bridge of approximately 307 feet in length and 52 feet in width. The bridge abutments and intermediate piers will be supported on a single row of drilled shafts. The finish grades will generally match existing grades. Besides fills for widening of the bridge approach, only minor site grading will be required for the project.

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FIGURE 2

Existing Bridge

1.2 Purpose and Scope

This report provides geotechnical recommendations based on engineering analyses of our subsurface investigation, including the general subsurface characterization, seismicity, boring logs, and laboratory test results. Geotechnical recommendations for the design of bridge abutments, wing walls, end posts, and site grading are presented in this report.

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2.0 Geology and Seismicity

2.1 Regional Geology

The Island of Kauai is composed of a single deeply eroded shield-shaped volcano with its associated intrusives, breccias, and tuff beds. These rocks and their remnants comprise the Waimea Canyon Volcanic Series, and were erupted during the late Tertiary Period. After the completion of the Kauai shield cone, there was a long period of erosion during which no volcanic activity occured. The return of volcanic activity resulted in new volcanic rocks spread over large areas of the eroded surface. The products of this period and renewed volcanic activity, along with closely associated sedimentary rocks, are known as the Koloa Volcanic Series.

The Waimea Canyon Series forms the main bulk of the Island of Kauai. The product of this series is divided into four distinct formations: Napali, Olokele, Haupu, and Makaweli. The formations consist of thin-bedded primitive-type olivine basalts (a`a), and massive olivine basalts (pahoehoe) that ponded in calderas and graben.

The eastern half of the Island of Kauai is overlain with lava flows of the Koloa Series. The lava flows consist of both pahoehoe and a`a formations extruded from groups of vents aligned in the north-south trends in various locales. The Koloa lava formation is underlain and interbedded with the Palikea formation, which consists of sedimentary breccias that grade laterally into stream-laid conglomerates resulting from the Waimea Canyon Series.

Changes in sea level during the Pleistocene Epoch caused the accumulation of deltas and fans of terrigenious sediments in the heads of Kauai's old bays, accumulation of reef deposits at correspondingly higher elevations, and lagoonal/marine sediments in the quiet waters protected by fringing reefs.

The project site is located in the Hanapepe area on the south side of the Island. The site lies within lavas of the Koloa Volcanic Series. Deposits of recent alluvium interbedded with beach and dune sand, as well as lagoonal clays and marls, are widespread along the project area. The recent alluvium is a result of material being transported to sea, and aggradation of the lower parts of valleys due to fluctuations in sea level.

Underlying the site are sedimentary breccias and conglomerates known as the Palikea Formation. At some places the conglomerates are associated with beds of sandstone and siltstone, largely of detrital origin, but in part tuffaceous.

2.2 Seismic Design Parameters

Based on the AASHTO site classification criteria, the predominant soils at the bridge site located at 21.908o

N latitude and -159.591o W longitude can be classified as a soft soil profile corresponding to a Seismic Site Class E.

Recommended seismic response parameters for use in design based on 7th Edition of the AASHTO LRFD Bridge Design Specifications represents ground motion corresponding to an exceedance probability of approximately 7 percent in 75 years for an earthquake with an approximate 1000-year return period. The 1000-year return period data for the soil at the bridge site was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short and long-period ground acceleration (SS and S1 respectively), and are summarized in Table 1.

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TABLE 1

Summary of Seismic Parameters for Earthquake with 1000-Year Return Period

Horizontal Peak Ground Acceleration, (PGA) 0.060g Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SS) 0.131g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (S1) 0.038g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 2.5 Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 2.5 Site Factor at Long-Period Range of Acceleration Spectrum, (FV) 3.5

Factored Horizontal Peak Ground Acceleration, (AS) 0.150g Factored Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDS) 0.327g Factored Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.132g

Seismic Zone Zone 1

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3.0 General Subsurface Characterization

Table 2 presents a summary of the subsurface characterization. Boring logs and laboratory test results are presented in the Appendixes A and B.

TABLE 2

Summary of Subsurface Materials for Foundation Design Purposes

Subsurface Depth

(ft) Material Description USCS

Classification Apparent Density

Ground Water Depth

(ft)

BORING HP1 (ELEVATION +12±) – WEST ABUTMENT, HANAPEPE BRIDGE

0.0 to 4.0 Silty GRAVEL with sand (Fill) GM Dense to medium dense

13.3

4.0 to 18.0 Clayey SILT MH Medium stiff to soft

18.0 to 29.0 Sandy SILT ML Soft

29.0 to 53.0 Clayey SILT with weathered sand MH Stiff

53.0 to 60.0 SAND, slightly cemented SP-SM Dense

60.0 to 85.5 Clayey SILT with weathered sand MH Stiff

BORING HP2A (ELEVATION +11.8±) – EAST ABUTMENT, HANAPEPE BRIDGE

0.0 to 14.5 Clayey SILT with gravel MH Medium stiff to soft

14.8

14.5 to 21.5 SAND with silt SP-SM Medium dense

21.5 to 26.0 Clayey SILT with sand and shell fragments ML Soft

26.0 to 99.5 Clayey SILT with weathered sand and cobbles MH Stiff

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TABLE 2 (continued) Summary of Subsurface Materials for Foundation Design Purposes

Subsurface Depth

(ft) Material Description USCS

Classification Apparent Density

Ground Water Depth

(ft)

BORING HP3 (ELEVATION -4.5±) – MIDDLE OF RIVER, HANAPEPE BRIDGE

0.0 to 4.0 SAND with silt and gravel SW-SM Loose

(-) 5.4

4.0 to 8.0 Clayey SILT with sand and shell fragments MH Soft

8.0 to 30.0 Silty SAND with gravel SM Medium dense

30.0 to 105.0 Clayey SILT with weathered sand and rock fragments MH Stiff

105 to 112 WEATHERED BASALT, highly to completely weathered -- Dense

112.0 to 124.0 COBBLES AND BOULDERS -- Dense

124.0 to 149.0 BASALT, moderately to highly weathered, interbedded with cobbles and boulders -- Medium hard to hard

BORING HP4 (ELEVATION +5.9±) – WEST APPROACH, HANAPEPE BRIDGE

0.0 to 4.0 Clayey SILT with sand MH Stiff

6.5

4.0 to 8.0 SAND with silt SP-SM Medium dense

8.0 to 17.0 Clayey SILT with sand ML Soft

17.0 to 20.0 Clayey SILT with sand MH Medium stiff

BORING HP5 (ELEVATION +10.6±) – WEST APPROACH, HANAPEPE BRIDGE

0.0 to 4.0 Silty GRAVEL GM Dense

10.8

4.0 to 10.0 Clayey SILT MH Stiff to soft

10.0 to 22.0 Silty SAND SM Medium dense to loose

22.0 to 24.5 Sandy SILT ML Soft

BORING HP6 (ELEVATION +9.5±) – EAST APPROACH, HANAPEPE BRIDGE

0.0 to 10.0 Clayey SILT with gravel MH Medium stiff to stiff 11.5

10.0 to 20.5 Silty SAND SM Medium dense to loose

The depth to groundwater can be expected to vary with tidal fluctuations and with water level within the Hanapepe River. A high groundwater level of elevation +2 is recommended for design.

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4.0 Analysis and Recommendations

This section provides a general interpretation of the site investigation and provides recommendations for design parameters at the proposed structure. Generalized subsurface profiles were developed based on data collected during the investigation. The primary structure considered for this project is replacing the existing three span bridge structure with a three span deep arch bridge structure.

4.1 Bridge Foundation

Recommendations are presented based on the use of 4-foot diameter drilled shafts. Based on discussions with the project structural engineer, we understand that four 4-foot diameter drilled shafts are being considered at the abutments and the piers.

The new bridge abutments will be located behind the existing abutments which will be removed. We understand that the roadway finish grade behind the abutments will generally match existing grades. As a result, global slope instability of the new abutments is not anticipated. In additional, since only minor additional fill is expected at the abutments, down drag due to settlements of new fill is also not anticipated.

Based on the factored ground acceleration coefficient at 1 second, SD1, the bridge site can be classified as Seismic Zone 1 in accordance to Table 3.10.6-1 of the AASHTO LRFD Bridge Design Specifications. In general, the potential of liquefaction in Seismic Zone 1 is low. Analysis was performed to determine the potential of liquefaction at the bridge abutments. Based on a factored horizontal peak ground acceleration of 0.15 for a 1,000-year return period, and a design earthquake magnitude of 6, factors of safety greater than 1.0 were computed. As a result, it is our opinion that the probability of potential soil liquefaction based on a seismic event with 1,000-year return period is low. Down drag from ground settlement due to liquefaction is therefore not anticipated.

Scour at the bridge abutments and piers will affect the axial and lateral resistance of the drilled shafts. Based on information provided on August 30, 2016, the potential scour at the proposed abutment and pier locations are as follows. Since scour resulting from incipient overtopping and pressure flood is more critical, it will be used for design at the abutments.

TABLE 3

Scour Depth and Elevation (8/30/16) 200-Year Event Total Scour (feet) Elevation (feet)

East abutment 2.37 -6.37 West abutment 2.37 -437

Piers 20.49 -25.49

500-Year Event Total Scour (feet) Elevation (feet)

East abutment 7.37 -11.37

West abutment 7.37 -9.37

Piers 24.01 -29.01

W.O. 15-5683.1 Page 8

TABLE 3 (continued) Scour Depth and Elevation (8/30/16) Incipient Overtopping and Pressure Flood Total Scour (feet) Elevation (feet) East abutment 8.83 -12.83

West abutment 8.83 -10.83

Piers 22.59 -27.59

Permanent casings are recommended for the upper 20 feet of the drilled shafts supporting the piers. The permanent steel casings should start from the bottom of the drilled shaft cap, extend to a depth of 20 feet.

In general, the permanent casing should have a nominal thickness of at least 0.5 inch.

4.1.1 Drilled Shaft Axial Capacity

We understand that the axial loads imposed on the drilled shafts will range from about 600 to 900 kips per drilled shaft at the abutments and 1250 to 1500 kips per shafts at the piers for Strength I Limit State. The drilled shafts will be spaced approximately 16 to 23 feet apart, measured from center to center, at the abutments and approximately 20 to 23 feet apart at the piers.

The drilled shafts will derive their axial load bearing capacity primarily from friction resistance developed between the perimeter of the shaft and the surrounding soils below the near surface soft clayey silt and loose silty sand. In general practice in Hawaii, the end-bearing component of the drilled shafts is discounted due to difficulties associated with obtaining a relatively clean bottom during construction of deep drilled shafts under groundwater. A generalized soil profile consisting of approximately 29 feet of medium stiff to soft clayey silt underlain by stiff clayey silt was used for drilled shaft design at the abutments. For drilled shafts at the piers, a generalized soil profile consisted of about 8 feet of soft clayey silt underlain by 22 feet of medium dense silty sand. The silty sand is underlain by stiff clayey silt and completely to moderately weathered basalt interbedded with cobbles and boulders at deeper depths. Assumed material properties for the various soil layers are summarized in Table 4. Unit side friction of the clayey silt material was determined by the α-method in Section 10.8.3.5.1b of the 7th edition of the AASHTO LRFD Bridge Design Specifications. Side resistance of the silty sand layer was determined by the β-method.

TABLE 4

Summary of Material Properties for Drilled Shaft Design

Shear Strength, Su (ksf)

Unit Side Friction, αSu (ksf)

Friction Angle

(degrees) Abutments Soft Clayey Silt 0.4 Disregarded -- Stiff Clayey Silt 4.0 2.05 -- Piers Soft clayey silt 0.4 Disregarded -- Medium dense silty sand -- -- 34 Stiff clayey silt 4.0 2.05 -- Completely to moderately weathered basalt and cobbles and boulders. -- 3.0 --

W.O. 15-5683.1 Page 9

We understand that Central Federal Lands Highway Division and Hawaii Department of Transportation have agreed that no load testing of the drilled shafts will be performed. Resistance factors are based on Table 10.5.5.2.4-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and a resistance factor of

0.50 is selected for strength limit state. Based on our analysis, the following drilled shaft lengths are recommended.

TABLE 5

Drilled Shaft Axial Resistance in Compression

Approximate Top of Drilled

Shaft Elevation (feet)

Drilled Shaft

Diameter

Drilled Shaft

Length

Strength Limit

State (kips)

Extreme Event

Limit State (kips)

Abutments -1 4 74 630 1,260 -1 4 90 900 1,800

Piers +2 4 139 1,360 2,720 +2 4 144 1,500 3,000

Structural loads at service limit state are not available at the time of this report. However, excessive settlement of drilled shafts is not anticipated.

4.1.2 Drilled Shaft Lateral Load Analysis

Lateral capacities of the drilled shafts will depend on the stiffness of the surrounding soil, the stiffness of the drilled shaft, the boundary condition at the top of the drilled shafts, the anticipated unsupported length at top, and the acceptable horizontal displacement of the shafts. Soil springs in the form of P-Y curves may be used to evaluate the lateral resistance of the drilled shafts. Subsurface soil parameters in Table 4 were used to develop the P-Y curves. Printout of P-Y curve data generated using LPILE (Ensoft, Inc.) is presented in Appendix C. The provided P-Y curves need to be adjusted to account for potential group effects. The selected group factors (P-Multiplier) should follow recommendations presented in Table 10.7.2.4-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications.

4.1.3 Drilled Shaft Construction

Based on our test borings, excavations for the drilled shafts can be expected to extend through loose to dense sand and silty sand, soft to stiff clayey silt, cemented sand and silt, gravels, cobbles, boulders, and basalt (pier locations). Due to the granular and cohesionless nature of the sand, silty sand, gravels, and cobbles, potential sloughing of the sidewalls of the drilled shaft excavation can be expected.

Rock drilling and coring equipment, as well as tools necessary for removal of the cored material, will be required for drilled shaft excavations extending into the hard basalt and cobbles and boulders layers.

We believe that either temporary casing or drilling slurry will be required during drilled shaft construction to prevent caving and sloughing of the drilled hole sidewalls. We recommend that temporary casing be used for construction of the drilled shafts. To facilitate advancement of the casing through the various soil layers that vary from soft/loose to stiff/dense to hard, the temporary casing should be equipped with cutting teeth and installed by rotating or oscillating methods.

W.O. 15-5683.1 Page 10

The bottom of the drilled hole should be cleaned prior to placement of concrete. The concrete should be placed as soon as practical upon completion of the drilled shaft excavations (within 24 hours). Concrete should be tremied through a pipe discharging below the surface of fresh concrete.

Construction of adjacent drilled shafts within three drilled shaft diameters should not commence until 24 hours after the concrete placement.

4.1.4 Trial Shafts and Load Test

We recommend that at least one trial shaft be constructed to determine the acceptability of the Contractor's equipment and procedures for drilled shaft construction, including concrete placement. The test shaft should be extended to the same depth as the production shafts. Once the test shaft is accepted, the same type of equipment and procedures demonstrated in the test shaft program should be used for construction of production shafts.

Load testing is usually performed to confirm the capability of the drilled shafts to support the design loads based on local practice. However, based on the small number of drilled shafts for this project, we understand that Federal Highway Administration’s Central Federal Lands Highway Division and Hawaii Department of Transportation have agreed no load test will be performed for this project.

4.1.5 Integrity Testing

Crosshole Sonic Logging (CSL) tests should be performed on all production drilled shafts as part of the quality control for drilled shaft construction. The downhole CSL method is a non-destructive integrity test that is based on the propagation of sound waves through concrete to assess the homogeneity of the drilled shafts, and to determine the location of anomalies, if any, in the concrete. The test should be performed in general accordance with ASTM D 6760.

To facilitate the CSL testing, access tubes should be embedded into the drilled shaft to allow the CSL probes, designed for receiving and transmitting ultrasonic waves, to enter the shaft. The recommended number of access tubes and tube spacing will depend on the drilled shaft diameter. The following are recommended number of access tubes for the 4-foot diameter drilled shafts.

TABLE 6

Recommended Number of Access Tubes

Drilled Shaft Diameter

Number of Tubes

Tube Spacing (deg)

Abutments and Piers 4 4 90

The access tubes should consist of standard steel pipe with an inside diameter between 1.5 to 2 inches extending from the bottom of the drilled shaft reinforcing cage to at least 3 feet above the top of the drilled shaft. The couplings and bottom cap of the access tubes should be watertight. The joints constructed along the full length of the access tubes should not hinder the passage of the CSL probes. The tubes should be filled with potable water prior to, or within one hour of, placing concrete in the drilled shaft. We also recommend that the top of the tubes be covered with removable caps to keep out debris which may obstruct the free passage of the CSL probes.

The CSL testing should be performed after the concrete of the drilled shaft has cured for at least 4 days.

Depending on the admixtures used in the concrete, a slightly longer cure time may be needed before testing. However, in order to reduce the potential for undesirable loss of ultrasonic energy due to de-

W.O. 15-5683.1 Page 11 bonding between the access tube and the surrounding concrete, we recommend that CSL tests be performed no later than 20 days after the concrete placement. The access tubes should be filled with non-shrink neat cement grout of the same strength as the drilled shaft after completion of the CSL tests.

In the event anomalies are detected by CSL testing, coring of the drilled shaft may be required to further evaluate the integrity of the concrete in the drilled shaft.

4.2 Abutment Walls and Wing Walls

Information regarding the wing walls was not available at the time of our report. Based on our borings behind the existing bridge abutments, the subsurface soil is relatively soft from depths below elevation +4.

As a result, we recommend that the wing walls be connected to the abutment walls/drilled shaft cap and supported by the drilled shafts. If needed, recommendations for smaller diameter drilled shafts can be provided. The abutment walls and wing walls should be designed to resist the lateral earth pressures presented in the Lateral Design section of this report.

4.3 Retaining Walls

Although finish grades were not available at the time of this report, we understand that retaining walls may be required behind the bridge abutments and wing walls. Based on existing topography, we assume that the walls will be relatively short, about 5 feet or less in height. Our borings drilled along the approaches to the bridge abutments generally encountered clayey silt soil in a medium stiff to stiff condition at near surface, transitioned to a softer condition at depths below 5 feet. Spread footings may be used to support the retaining walls. To provide more uniform support, we recommend that all footings be underlain by a minimum 12 inches of aggregate base course. The base course should be compacted to a minimum 95 percent compaction as determined by AASHTO T-180 (ASTM D 1557). The subgrade soil should be compacted to a minimum 90 percent compaction. Soft or loose soils, indicated by pumping conditions, should be removed and replaced with aggregate base course or imported granular structural fill. Global slope instability of the 5 feet high retaining walls are not anticipated.

4.3.1 Spread Footing Bearing Resistance

Based on our borings, the near surface clayey silt is in a medium stiff to stiff condition. The following bearing values are recommended for design of the footings. The bearing resistance factors are based on Table 10.5.5.2.2-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications, and the applied strength limit state resistance factor for bearing resistance is 0.45. An allowable bearing resistance of 2,000 psf is recommended for service limit state in order to limit the potential settlement to less than 1-inch.

TABLE 7

Bearing Value for Retaining Wall Foundations

Extreme Event Limit State 6,000 psf

Strength Limit State 2,700 psf

Service Limit State 2,000 psf

Spread footings should be a minimum of 24 inches in width. The footings should be kept as high as possible but embedded at least 12 inches below finish adjacent grade. Footings located on or near the top of slopes should be embedded such that a minimum horizontal distance of 6 feet is maintained between the bottom

W.O. 15-5683.1 Page 12 edge of footing and slope face. The bottom of footing excavations should be thoroughly cleaned of loose material prior to placement of reinforcing steel and concrete.

4.3.2 Footing Lateral Resistance

Resistance to lateral loading may be provided by friction acting at the base of foundations and by passive earth pressure acting on the buried portions of foundations. The following coefficients of friction are recommended. Resistance factors are based on Table 10.5.5.2.2-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications. The applied strength limit state resistance factor for coefficient of friction is

0.80. Passive earth pressures are presented in the Lateral Design section of this report.

TABLE 8

Coefficient of Friction

Coefficient of Friction for Extreme Limit State 0.67

Coefficient of Friction for Strength Limit State 0.54

4.4 Lateral Design

Resistance to lateral loading may be provided by passive earth pressure acting on the buried portions of the foundations. The following soil parameters were used to determine the earth pressures for design.

TABLE 9

Summary of Material Property for Design

Material Description Unit Weight (pcf)

Cohesion (psf)

Friction Angle (deg)

Structural Fill 130 0 34

Onsite Clayey Silt 115 200 30

4.4.1 Passive Earth Pressures

Passive earth pressures were computed using the Log-Spiral Method. A wall friction angle value equal of 1/2 times the soil’s friction angle was used. Based on the computed values and local practice, the following passive earth pressures are recommended. Resistance factors are based on Table 10.5.5.2.2-1 in the 7th Edition of the AASHTO LRFD Bridge Design Specifications. The applied strength limit state resistance factor for passive pressure resistance is 0.5. Unless covered by pavement or concrete slabs, the upper 12 inches of soil should not be considered in computing lateral resistance.

TABLE 10

Passive Pressure

Soil Type Extreme Event Limit State (pcf)

Strength Limit State (pcf)

Onsite Clayey Silt 500 250

4.4.2 Active and At-rest Earth Pressures

Active earth pressures were computed using the Coulomb Criteria. Based on the computed values and local practice, the following active earth pressures are recommended. In general, a movement of as much as 0.5

W.O. 15-5683.1 Page 13 percent of the wall height at the top of wall by rotation or lateral translation may be required to reach the active pressure condition. Walls that are not free to deflect sufficiently to permit development of active earth pressure should be designed using the at-rest condition.

TABLE 11

Active / At-rest Pressures

Material Description

Level Backfill

Condition (pcf)

Maximum 2H:1V Sloping Backfill

Condition (pcf)

Restrained / At-Rest

Level Condition (pcf)

Structural Fill 33 52 57

Onsite Clayey Silt 35 60 58

4.4.3 Dynamic Lateral Earth Forces

For dynamic lateral earth pressure considerations, the following total dynamic lateral forces may be assumed for seismic loading with maximum peak ground acceleration (amax) of 0.15g. The recommended dynamic lateral force for unrestrained condition assumes that the wall is free to move laterally up to 1 to 2 inches or rotate in the event of an earthquake. The dynamic lateral force may be assumed to act through the mid-height of the wall.

TABLE 12

Dynamic Lateral Earth Forces (Level Backfill Condition)

Wall Height, H (feet)

Backfill Material

Un-Restrained Condition

Restrained Condition

4 Structural Fill 7.2H2 10.1H2

6 Structural Fill 11.3H2 14.3H2

8 Structural Fill 13.4H2 16.5H2

13 Structural Fill 15.8H2 19.0H2 Note: The dynamic lateral earth force has a unit of pounds per lineal foot of wall length.

The above dynamic lateral earth forces included the static earth pressures and were based on procedures in Section A11.3.2 of the 7th Edition of the AASHTO LRFD Bridge Design Specifications. For un-restrained condition, kh equals to 1/2 times the maximum peak ground acceleration was used. For restrained condition, kh equals to maximum peak ground acceleration.

4.4.4 Drainage

To prevent buildup of hydrostatic pressures, retaining walls should be well-drained. The standard of practice consists of placing a minimum 12-inch thick layer of free-draining gravel at the back of the wall. The gravel should extend from the base of the wall, around subdrains and/or weepholes, and up to within 12 inches of finish grade.

Alternatively, prefabricated drainage geocomposites, such as Miradrain or J-drain, may be used in lieu of the free-draining gravel. As with the free-draining gravel, the drainage geocomposites should be placed at the back of the wall, be connected with the weepholes and/or subdrains (in accordance with manufacturers specifications), and extend to within 12 inches of finish grade. For freestanding walls, the drainage system

W.O. 15-5683.1 Page 14 should be covered by at least 12 inches of low permeability soil, such as silty clay or clayey silt with low expansion potential.

4.5 Bridge Approach Slab

We understand that approach slabs will be constructed behind the bridge abutments. The slabs should be underlain by at least 8 inches of aggregate base course. The base course and subgrade should be compacted to a minimum 95 percent compaction as determined by AASHTO T-180 (ASTM D 1557). The subgrade soil should be compacted to a minimum 90 percent compaction.

4.6 Pavement Section

The AC concrete pavement encountered in borings HP1, HP2A, and HP5 consisted of approximately 3 inches of asphaltic concrete over approximately 4 to 9 inches of base material.

Due to the relatively short length of pavement for bridge approaches, based on discussions with Central Federal Lands Highway Division, we understand that new pavement will generally match the existing pavement sections identified in the borings or thicker.

4.7 Site Grading

4.7.1 Site Preparation

The project site should be cleared of all vegetation, AC pavements, and other deleterious material. Prior to fill placement, the existing ground should first be scarified to a depth of six inches, moisture conditioned to about 2 percent above optimum moisture content, and compacted to a minimum 90 percent compaction as determined by AASHTO T-180 (ASTM D 1557). Soft or loose soils, indicated by pumping conditions, should be removed and replaced with imported granular structural fill.

4.7.2 Structural Excavations

Based on our exploratory test borings, we believe that excavations into the surface clayey silt and silty gravel can be accomplished with conventional excavating equipment. Temporary cuts should be stable at slope gradients of 1H:1V (horizontal to vertical) or flatter for temporary conditions. Shoring may be required for excavations extending into the soft/loose deposits below groundwater. It should be the Contractor’s responsibility to conform to all OSHA safety standards for excavations.

4.7.3 Embankments

Widening of the bridge approaches will require fill placement on the sides of the existing Kaumualii Highway behind the bridge abutments. The width of the fill is expected to be about 5 feet, with fill thickness on the order of 2 to 5 feet. Maximum vertical height of the slope is about 7 feet. Ground settlement due to the new fills is expected to be less than one inch. Fill slopes should be constructed at slope gradients of 2H:1V or flatter as recommended in the following section. Global slope instability for the 2H:1V slopes is not anticipated.

4.7.4 Slope Gradients

All cut and fill slopes should be stable at gradients of 2H:1V or flatter. All slopes should be planted as soon as practical to reduce the effects of erosion and weathering.

W.O. 15-5683.1 Page 15

Fill slopes should be constructed from the bottom up. Fill placed on slopes should be keyed and benched into the existing slope to provide stability for the new fill against sliding. Filling the slope with sliver fills should be avoided. Fill placed in areas which slope steeper than 5H:1V should be continually benched as the fill is brought up in lifts. Fill slopes should be constructed by overfilling and cutting back to the design slope gradient to obtain a well-compacted slope face.

4.7.5 Onsite Fill Material

The onsite clayey silt and silty gravel above groundwater table may be reused in compacted fills and backfills, provided all rock fragments larger than 3 inches in maximum dimension are removed. In addition, the moisture content of the clayey silt should be maintained at about 2 percent above the optimum moisture content during recompaction.

4.7.6 Imported Fill Material

Imported structural fill should be well-graded, non-expansive granular material and conform to FP-14 704.04 with modifications presented in 5.0 Special Contract Requirements section of this report.

4.7.7 Backfill for Bridge Abutment Walls and Wing Walls

Backfill for the retaining structures, including abutment walls and wing walls, should consist of granular structural fill such as the imported structural fill material.

4.7.8 Compaction

Structural fill, granular backfill, and granular embankment fill should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 95 percent compaction as determined by AASHTO T-180 (ASTM D 1557). Fill consisting of cohesive soils such as the onsite clayey silt should be placed in horizontal lifts restricted to 8 inches in loose thickness and compacted to a minimum 90 percent compaction.

Fill placed in areas which slope steeper than 5H:1V should be continually benched as the fill is brought up in lifts. Fill placed on slopes should be keyed and benched into the existing slope to provide stability for the new fill against sliding.

W.O. 15-5683.1 Page 16

5.0 Special Contract Requirements

Section 704.04 Structural Backfill. Furnish a suitable, well-graded, non-expansive granular material conforming to the following:

(a) Maximum particle size 3 in (75 mm)

(b) Material passing No. 200 (75-µm) between 8 and 20 percent sieve, AASHTO T 27 and AASHTO T 11

(c) Plasticity index, AASHTO T 90 10 max.

(d) California Bearing Ratio, AASHTO T 193 CBR value, 15 min.

(ASTM D 1883) CBR expansion value, 1.0 percent max.

W.O. 15-5683.1 Page 17

6.0 Additional Services

We recommend that we perform a general review of the final design plans and specifications. This will allow us to verify that the foundation design and earthwork recommendations have been properly interpreted and implemented in the design plans and construction specifications.

For continuity, we recommend that we be retained during construction to (1) observe all drilled shaft construction work, including trial shafts and load test, (2) observe footing excavations prior to placement of aggregate base course layer, reinforcing steel and concrete, (3) review and/or perform laboratory testing on import borrow to determine its acceptability for use in compacted fills, (4) observe structural fill placement and perform compaction testing, and (5) provide geotechnical consultation as required.

Our services during construction will allow us to verify that our recommendations are properly interpreted and included in construction, and if necessary, to make modifications to those recommendations, thereby reducing construction delays in the event subsurface conditions differ from those anticipated.

W.O. 15-5683.1 Page 18

7.0 Limitations

The boring logs indicate the approximate subsurface soil conditions encountered only at those times and locations where our borings were made, and may not represent conditions at other times and locations.

This report was prepared specifically for CH2M Hill and their sub-consultants for design of the proposed replacement of Hanapepe River Bridge in Kauai, Hawaii. The boring logs, laboratory test results, and recommendations presented in this report are for design purposes only, and are not intended for use in developing cost estimates by the contractor.

During construction, should subsurface conditions differ from those encountered in our borings, we should be advised immediately in order to re-evaluate our recommendations, and to revise or verify them in writing before proceeding with construction.

Our recommendations and conclusions are based upon the site materials observed, the preliminary design information made available, the data obtained from our site exploration, our engineering analyses, and our experience and engineering judgment. The conclusions and recommendations in this report are professional opinions which we have strived to develop in a manner consistent with that level of care, skill, and competence ordinarily exercised by members of the profession in good standing, currently practicing under similar conditions in the same locality. We will be responsible for those recommendations and conclusions, but will not be responsible for the interpretation by others of the information developed. No warranty is made regarding the services performed, either expressed or implied.

Con C. Truong, Project Engineer

Rick Yoshida, Project Manager

This work was prepared by me or under my supervision.

Expiration Date of License:

April 30, 2018

Appendix A Field Exploration Program & Boring Logs

Plate A1.1

Appendix A: Field Exploration Program & Boring Logs

A1.0 Introduction Hirata & Associates, Inc. completed a field exploration program for Hanapepe River Bridge, Kaumualii Highway, Route 50, District of Waimea, Island of Kauai, State of Hawaii, Task Number 499067.09.SU.HI, between October 22 2014, and February 6, 2015, for the new bridge. The scope of work for the field exploration program included three borings for the new bridge and three borings for proposed retaining walls along the bridge approaches. Borings for the proposed detour bridge were eliminated after the location of the detour bridge was revised. Each component of the field exploration program was coordinated and observed by a Geotechnical Engineer from Hirata & Associates, Inc. Boring location plans, as well as individual boring logs are attached. These logs represent a compilation of field data and description of the soil samples by Hirata & Associates, Inc. personnel. The methods used to conduct the field exploration program are described below. Representative soil and rock samples collected during the field exploration program were transported to Hirata & Associates, Inc. in Aiea, Oahu, Hawaii for testing. A summary of the laboratory testing program is provided in Appendix B.

A2.0 Explorations Hirata & Associates, Inc. of Aiea, Oahu, Hawaii provided the drilling services for the soil borings. Borings were completed using a Mobile B80 truck-mounted drill rig and portable drilling equipment.

The initial phase of borings for the new bridge was drilled between October 22 and November 10, 2014.

Two borings at the rear of existing bridge abutments, HP1 and HP2A, were completed to depths of about

85.5 and 99.5 feet below ground surface to explore subgrade conditions. The borings for the proposed retaining walls along the bridge approaches were drilled on November 12 and 13, 2014. Three borings along the existing bridge approaches, HP4, HP5, and HP6, were completed to depths ranging from about 20.5 to

24.5 feet below ground surface to explore subgrade conditions. One boring within the Hanapepe River adjacent to the existing bridge was drilled between January 29 and February 6, 2015. The boring, HP3, was completed to a depth of about 149 feet below mud line. Borings were drilled and sampled using continuous flight solid stem augers and continuous sample rock coring techniques.

The locations of individual borings were located in the field by measuring/taping offsets from existing features shown on the plans. Boring locations are shown on the Boring Location Plan, Plate A2.2, and are shown below in Table A-1. Surface elevations at boring locations were estimated based on the Topographic Survey provided by CH2M Hill received on December 22, 2014. Following drilling activities, the borings were backfilled with cuttings generated during the drilling operations and borings HP1, HP2A, and HP5 were capped with “Instant Road Repair” by International Roadway Research.

Plate A1.2

TABLE A-1

Field Exploration Locations

Foundation Designation Boring

Number Station Offset Ground

Elevation Termination

Elevation West Abutment HP1 205+64.2 5.9 RT +12 ft -73.5 ft East Abutment HP2A 208+80.9 6.1 RT +11.8 ft -87.7 ft

Piers HP3 207+30.5 22 LT -4.5 ft -153.5 ft West Approach HP4 204+45.8 33.1 LT +5.9 ft -14.1 ft West Approach HP5 204+82.6 13 RT +10.6 ft -13.9 ft East Approach HP6 209+67.3 28.1 LT +9.5 ft -11.0 ft

A3.0 Soil and Rock Sampling Disturbed samples were obtained from the borings in general accordance with ASTM D 1586 for the Standard Penetration Test (SPT). The SPT involves driving a 2-inch outside diameter, 1.375-inch inside diameter split-spoon sampler a depth of 24 inches with a 140-pound hammer falling a distance of 30 inches.

The number of blows required to advance the split-spoon sampler through each of the 6-inch increments were recorded. The SPT resistance, or N-value, is defined as the number of blows required to drive the sampler over the second and third 6-inch increments. The N-value provides a means for evaluating the relative density or compactness of cohesionless (granular) soils and consistency or stiffness of cohesive (fine-grained) soils. Because high penetration resistance prevented driving the total length of the sampler at times, the penetration resistance for the partial penetration was recorded where applicable. Representative portions of the split-spoon sample obtained in conjunction with the SPT were placed in plastic bags and transported to the Hirata & Associates, Inc. for testing.

Samples were also obtained using a modified California (MC) barrel sampler in general accordance with ASTM D 3550. The MC test procedure is similar to the SPT, except a large diameter barrel sampler (3-inch O.D., lined with 2.42-inch-diamter brass sleeves) is used and driven 18 inches. As a result of the larger diameter, the MC sampler yields slightly higher raw blow counts when compared to SPT N-values for similar soils. The raw MC blow counts over the second and third 6-inch increments were used to define the relative density and consistency of the subsurface materials. MC samples were sealed with plastic end caps and transported to Hirata & Associates, Inc. for testing.

Rock samples were collected from the borings using a NX core barrel having an inside diameter of 2.1 inches.

During coring, cuttings were removed from the borehole by circulating water down the drill rods and back up the annulus of the boring. Core runs ranged between approximately 4 and 5 feet in length. Percent recovery and rock quality designation (RQD) values are shown on the boring logs. Percent recovery is the ratio of the length of recovered core to the total length of the core run. RQD is the ratio of the sum of the length of intact core pieces greater than 4 inches long in a run to the total length of the core run. Following drilling, the cores were logged by Hirata & Associates, Inc. personnel, then stored in cardboard core boxes and transported to the Hirata & Associates, Inc. laboratory.

A4.0 Soil and Rock Classification System During the drilling of borings, Hirata & Associates, Inc. personnel collected soil/rock samples and prepared field logs of the borings. Soil identification and descriptions, as shown on the field logs, are based on ASTM D2488, a systematic process for identifying and describing individual soil samples by visual and manual means. When sufficient laboratory testing was completed, select samples from borings were classified using

Plate A1.3 the Unified Soil Classification System (USCS). Both the visual soil identification system and the referenced soil classification systems are summarized in the attached Boring Log Legend and USCS plates. Rock samples were classified based on the stratigraphic structure, rock strength, degree of weathering, and other properties. The rock classification system is summarized in the attached Rock Weathering Classification System.

Appendix B Laboratory Test Results

Plate B1.1

Appendix B: Laboratory Test Results

B1.0 Introduction Laboratory tests were completed on select soil samples recovered from the field exploration program in general accordance with ASTM testing methods. The laboratory testing program was completed to provide data for engineering studies and to classify the materials into similar geologic groups. The testing program included index tests and geotechnical engineering property tests. The following sections describe the laboratory testing procedures and test results are presented in the attached laboratory reports.

B2.0 Index Tests Classification and index laboratory testing included identification by visual and manual means, and tests to determine natural water content, unit weight, grain size distribution, and Atterberg limits. When sufficient laboratory testing was completed, select samples from borings were classified using the Unified Soil Classification System (USCS). Both the visual soil identification system and the referenced soil classification system are summarized in the Boring Log Legend and USCS plates in Appendix A. The final classifications are shown at the appropriate locations on the Boring Logs, Plates A4.1 through A4.15. Index tests are generally conducted on disturbed soil samples. The following sections describe individual index test procedures.

B2.1 Moisture Content Water content was determined for samples retrieved from the exploration in general accordance with ASTM D 2937. To perform this test method, the sample was weighed before and after oven drying, and the water content was calculated. The moisture content of soils, when combined with data obtained from other tests, provides information about the characteristics of the soil, including general correlations with strength, settlement, and workability.

B2.3 Atterberg Limits Liquid and plastic limit tests were performed on selected fine-grained samples. The tests were completed in general accordance with ASTM D 4318. The Atterberg limits include liquid limit (LL), plastic limit (PL), and plasticity index (PI), which is the plastic limit subtracted from the liquid limit. The results of these tests are presented on the USCS plate in Appendix A, and in the following table. These limits are generally used to assist in classification of soils, to indicate soil consistency, and to provide correlation to engineering properties.

TABLE B-1

Atterberg Limit Test Results

Sample Liquid Limit (LL) Plastic Limit (PL) Plasticity Index (PI) HP2A at 3.5 feet 53 34 19 HP3 at 24 feet 71 57 14 HP3 at 74 feet 62 41 21 HP5 at 4 feet 53 33 20

Plate B1.2

B2.4 Gradation The grain size distribution of selected samples was determined in general accordance with ASTM D 422.

The results of these tests are presented on Plates B4.1 and B4.2. These tests aid in the classification of soils and provide correlating data with engineering properties of soils, such as permeability, strength, settlement, and swelling potential.

B3.0 Geotechnical Engineering Property Tests For Soil Geotechnical engineering property testing for soil included In-place Density, Triaxial Strength, Direct Shear and Consolidation tests. Geotechnical engineering property test results are presented in the attached laboratory reports. The following sections describe these test procedures for soil.

B3.1 In-place Density In-place density tests of select, undisturbed samples were performed in general accordance with ASTM D 2937. This test is performed on relatively undisturbed samples recovered from drive cylinders. The results of this test are shown at the appropriate depths on the Boring Logs, Plates A4.1 through A4.15, and are used to determine the in-place density of soils.

B3.2 Consolidation Consolidation tests were performed on selected relatively undisturbed samples in general accordance with ASTM D 2435 to determine the compression characteristics, creep, stiffness, and flow properties of soils under loading. The samples were inundated during testing to represent adverse field conditions.

The percent of consolidation for each load cycle was recorded as a ratio of the amount of vertical compression to the original height of the sample. The results of the tests are presented on Plates B2.1 through B2.4 as a graph showing effective vertical stress versus percent change in height and are critical in completing settlement analyses.

B3.3 Direct Shear Direct shear tests were performed on undisturbed samples in general accordance with ASTM D 3080 to determine the shear strength characteristics of selected materials along a predetermined horizontal plane. The results of these tests are presented on Plates B3.1 and B3.2 as graphs showing shear stress versus normal stress and are used in a variety of geotechnical engineering analyses, including slope stability, foundation design, and earth retention design.

B3.5 Triaxial Strength Unconsolidated, undrained triaxial strength tests were completed on a selected, undisturbed sample in general accordance with ASTM D 2850 to determine the…