FINAL_GEOTECH_RPT_CA_BLM_19041.pdf

GUIDOTTI BRIDGE

FORT ORD NATIONAL MONUMENT

CA BLM 19041 (2)

FINAL GEOTECHNICAL REPORT

Report # CA-BLM-0001-14-01

Geotechnical Section Technical Services Branch

January 2015 i | P a g e Guidotti Bridge, Fort Ord National Monument Project CA BLM 19041 (2)

Table of Contents

SECTION ONE--INTRODUCTION

1.1 OBJECTIVE AND SCOPE

SECTION TWO--GEOLOGY AND SEISMICITY

2.1 GEOLOGY

2.2 GEOLOGIC HAZARDS

2.3 SEISMICITY

2.4 SEISMIC DESIGN PARAMETERS

SECTION THREE--SUBSURFACE INVESTIGATION

3.1 FIELD EXPLORATION

3.2 LABORATORY TESTING

SECTION FOUR--ANALYSIS AND RECOMMENDATIONS

4.1 BRIDGE FOUNDATION DESIGN

4.1.1 Geotechnical Profile

4.1.2 Groundwater Conditions

4.1.3 Liquefaction Potential

4.1.4 Liquefaction Induced Settlement

4.1.5 Pre- and Post-Liquefaction Properties for Lateral Resistance

4.1.6 Bridge Foundation Selection

4.1.7 LRFD Driven Pile Foundation Design

4.1.8 Downdrag Load

4.1.9 Nominal Axial Resistance

4.1.10 Pile Setup

4.1.11 Group Effects on Axial Resistance

4.1.12 Lateral Loads

4.1.13 Pile Driveability

4.1.14 Settlement

4.1.15 Field Testing

4.2 ABUTMENT DESIGN

4.3 CORROSION POTENTIAL

4.4 SHRINK/SWELL RECOMMENDATIONS

4.5 CONSTRUCTION CONSIDERATIONS

4.6 SPECIFICATIONS

4.7 LIMITATIONS

SECTION FIVE -- REFERENCES

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TABLES

TABLE 1: Fault Data TABLE 2: Summary of Seismic Parameters Corrected for Site Class D TABLE 3: Summary of Design Response Spectrum Data for Site Class D Soil TABLE 4: Summary of Bridge Borings TABLE 5: Summary of Soil Classification Index Tests TABLE 6: Summary of Electrochemical Properties TABLE 7: Proposed Bridge Foundation Locations TABLE 8: N1(60) Calculations for Overburden Material at Bridge Location TABLE 9: Estimated Subsurface Material Properties for Bridge Foundation Design TABLE 10: Soil Characterization for Pre- and Post-Liquefaction TABLE 11: Summary of Nominal Pile Resistance (Target Resistance) per Pile TABLE 12: Summary of Pile Recommendations TABLE 13: Lateral Earth Pressures for Bridge Abutments

ILLUSTRATIONS

ILLUSTRATION 1: Design Response Spectrum Corrected for Class D Soil ILLUSTRATION 2: Histogram for the Deaggregation Hazard Illustrating the Potential

Source Distances from the Bridge Site

FIGURES

FIGURE F-1: Bridge Location Map FIGURE F-2: Plan and Elevation Sheet FIGURE F-3: Foundation Plan Sheet (100% plans) FIGURE F-4: Boring Log Plan Sheet (100% plans) FIGURE F-5: Slope Protection Plan Sheet (100% plans)

APPENDICES

APPENDIX A - Field Investigation and Boring Logs APPENDIX B - Laboratory Test Results APPENDIX C - Photographs APPENDIX D- APILE Analysis APPENDIX E- Pile Driving Analysis and Potential Liquefaction Analysis APPENDIX F- Special Contract Requirements

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SECTION ONE--INTRODUCTION

This report presents the results of the geotechnical engineering investigation conducted for the Guidotti Pedestrian Bridge replacement, CA BLM 1941 (2) project. The bridge site is located within Fort Ord National Monument on Guidotti Road between Monterey and Salinas in Monterey County, CA. This project is a coordinated effort between the Bureau of Land Management (BLM), and the Federal Highway Administration- Central Federal Lands Highway Division (FHWA-CFLHD) to replace the existing structurally deficient bridge with a structure capable carrying heavy construction loads. The existing gravel roadway width generally varies from 12 to 14 feet, and is generally used by light vehicles and pedestrians. No roadway geometric improvements to Guidotti Road are anticipated beyond minor adjustments required near the bridge approaches.

The existing Guidotti Bridge is a three-span filled concrete spandrel arch crossing over the El Toro Creek. The middle section of the arch was constructed in 1908 and the two exterior spans were constructed in 1911 as noted on dates embossed on bridge. The existing bridge has failed due to scour and subsequent settlement of spread footings. Bridge inspection report also indicated that seismic loading may have been a contributing factor to bridge failure.

The Guidotti Bridge is currently a one-lane crossing mainly used by pedestrians, cyclists, and equestrians to access recreation trails of Fort Ord National Monument. BLM desires replacement structure to provide vehicular access for heavy construction and fire equipment.

The new bridge will be located at same location on existing straight alignment to minimize environmental impacts. The bridge is anticipated to be completely replaced within relatively short period and hence no temporary access will be required during construction. The new bridge will be designed to meet current AASHTO LRFD bridge design standards for HL93 live Load limits.

1.1 OBJECTIVE AND SCOPE

The objectives of this investigation were to develop recommendations concerning structure foundations, subsurface drainage, corrosion potential, and shrink/swell values. In accomplishing these objectives, the scope of work included field exploration, laboratory testing, correlation of available data, and engineering analysis and recommendations. The field investigations included subsurface exploration conducted by CFLHD engineering geologist. The interpretation of laboratory testing, correlation of available data, and engineering analysis was conducted by geotechnical personnel from CFLHD. This report presents the findings of the subsurface investigation programs and provides geotechnical recommendations for the design and construction of the Guidotti Bridge. Foundation design recommendations included in this report are consistent with AASHTO LRFD and local design guidelines. The report also provides recommended Special Contract Requirements.

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SECTION TWO--GEOLOGY AND SEISMICITY

2.1 GEOLOGY

REGIONAL GEOLOGY

Fort Ord National Monument is located within the California Coast Ranges Section of the Pacific Border physiographic province. The Coast Ranges are characterized by a linear system of nearly parallel ranges and are dominated by active tectonics at the margin of the Pacific and North American tectonic plates. These ranges are generally geosynclinal in structure, with subsequent folding and faulting in late Pliocene. The landscape is moderately dissected rolling hills underlain by young terrestrial deposits.

The region is bordered by outcrops of Pleistocene and recent coastal beach and dune deposits which are bordered by Pleistocene stabilized dunes inland. The stabilized dunes cover roughly one-half the area of Fort Ord. The Aromas Sandstone occasionally outcrops within the stabilized dunes area, as well as underlies the stabilized dunes. The Aromas Sandstone is underlain by the Paso Robles Formation and the Santa Margarita Sandstone.

The Quaternary Aromas Sandstone consists of a relatively clean red to yellow brown, well sorted sand with silt and clay lenses, and ranges in thickness from 50 feet near the coast to up to 200 feet inland. The Plio-Pleistocene Paso Robles Formation consists of sand, gravel, and clay interbedded with minor calcareous beds. The Upper Tertiary Santa Margarita Sandstone consists of a poorly consolidated marine sandstone.

LOCAL GEOLOGIC CONDITIONS

The soils at Guidotti Bridge along Toro Creek and sporadically throughout the area are comprised of alluvial soils consisting of sands overlying silt and clay of alluvial and eolian (dune) origin. At boring B-1 or the abutment 1 location, well-graded silty sand overlying silty sand with varying amounts of gravel, and sandy silt layers were encountered. Standard penetration test (SPT) blow count (N) values ranged from 7 to refusal (50 blows for less than a 6–inch depth) and generally increased with depth. At boring B-2 or the abutment 2 location, silty sand overlying gravelly sand, clay, and silt layers were encountered. SPT N values ranged from 9 to refusal and generally increased with depth. Groundwater was encountered at both boring locations, at depths ranging from 13 to 13.5 feet.

Boulders up to 1 foot diameter are present on the surface in the vicinity of the bridge site and should be anticipated during construction.

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2.2 GEOLOGIC HAZARDS

Geologic hazards exist from both the natural environment of the project site and from existing and proposed construction of the bridge site. Geologic hazards that exist in the vicinity of the project area are seismic events (detailed in the following Sections 2.3-Seismicity and 2.4-Seismic Design Parameters), flash floods, highly erodible soils, and slope instability at fill slope locations. Flash floods are possible in the creek drainage area during periods of high precipitation. Highly erodible fill slopes along the embankment fills leading up to the bridge are susceptible to surface water erosion, which means directing surface water flow and properly constructing riprap protection for the embankments as shown in the plans is important for this site.

2.3 SEISMICITY

The closest active faults to the project are listed in Table 1, Fault Data. The San Andreas Fault, which has been historically active within the last 200 years, is within 25 miles of Fort Ord. The potential of earthquake damage from ground movement is moderate to very high, with the highest potential in the alluvial sediments that exist along Toro Creek.

TABLE 1: Fault Data

Fault Name

Maximum Moment

Magnitude (Mmax)

Style of Faulting

Fault Dip, degrees

Fault Strike

Length, miles

Distance, miles

Reliz Fault Zone 6.3 Reverse 58W NW 48.1 4.7 Monterey Bay - Tularcitos 7.3 Dextral 90 NW 52.5 4.9 San Gregorio Fault Zone 7.5 Dextral 90 NW 150.6 14.2 San Andreas (Creeping Section) 7.7 Dextral 90 NW 676.3 20.4 Calaveras (so) – Paicines Extention 6.9 Dextral 77E NW 97.5 24.3 Sargent Fault Zone 6.4 Dextral 90 NW 35.0 24.7 Quien Sabe 6.4 Dextral 90 NW 15.0 28.3

2.4 SEISMIC DESIGN PARAMETERS

The American Association of State Highway and Transportation Officials (AASHTO) criteria for seismic design of ordinary bridges, allows for damage to occur when the structure is subjected to 1000-year strong ground motion, but does not allow for collapse of one or more spans and loss of life. Bridges may suffer damage that requires repair or partial replacement.

However, AASHTO requires bridges to withstand small, more frequent earthquakes without damage. For Extreme Event I Limit State Design an earthquake load and resistance factor of 1.00 should be used during the LRFD design of the bridge structure.

Recommended seismic response parameters for the Guidotti Bridge site design are based on the (AASHTO) LRFD (2012), and represents horizontal peak ground acceleration (PGA) with 7 percent probability of exceedance in 75 years (approximately 1000-year return period). The

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1000-year return period uniform hazard spectrum for the bridge site, located at 36.5831º N latitude and -121.7164º W longitudinal, was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short- and long- period ground acceleration (Ss and S1 respectively) and corrected for the soil profile at the bridge site.

Ground motions caused by earthquakes are influenced not only by the distance from the fault planes, but also by the geology and soils found at the site. Subsurface profiles similar to the ones at the project site, with soil layers overlaying bedrock, will have amplified ground motions and a resonant period governed by the layer thickness and shear wave velocity of the materials. The damage potential of strong ground motions with respect to a structure is typically affected by the period of the strong earthquake motion and the resonant period of both the soil and the structure.

Based on the subsurface profile at the bridge site, the average time-weighted shear wave velocity for the top 100 feet (VS100) of subsurface materials was estimated between 600 and 1,200 feet per second based on observed average Standard Penetration Test (SPT) blow counts between 15 and 50 blows per foot. Therefore, the site was classified as Site Class D according to the site class definitions specified in Table 3.10.3.1-1 of ASSHTO.

A seismic hazard analysis to establish ground motions for seismic design was conducted. The recommended spectral acceleration coefficient values for probabilistic design with a return period of 1000 years were calculated using the program provided with the AASHTO LRFD Bridge Design Manual developed by the USGS (2008) entitled “Seismic Design Parameters”, version 2.10 and are summarized in Table 2, Summary of Seismic Parameters Corrected for Site Class D Soils.

TABLE 2: Summary of Seismic Parameters Corrected for Site Class D

Horizontal Peak Ground Acceleration, (PGA) 0.381g Horizontal Response Spectral Acceleration at Period of 0.2 sec, (Ss) 0.904g Horizontal Response Spectral Acceleration at Period of 1.0 sec, (S1) 0.394g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.12 Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.14 Site Factor at Long-Period Range of Acceleration Spectrum, (Fv) 1.61

Factored Horizontal Peak Ground Acceleration, (As) 0.427g Factored Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDs) 1.031g Factored Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.634g

The 5% damped design response spectrum for Site Class D is shown in Illustration 1, Design Response Spectrum Corrected for Site Class D, and the supporting data is presented in Table 3, Summary of Design Response Spectrum Data for Site Class D.

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ILLUSTRATION 1:- Design Response Spectrum Corrected for Class D Soil

TABLE 3: Summary of Design Response Spectrum Data for Site Class D Soil

Period (sec) Sa(g) Sd (in)

0.000 0.427 0.000 (PGA)(Fpga) = As

0.123 1.031 0.153

0.200 1.031 0.403 (Ss)(Fa) = SDs

0.526 1.031 3.814

0.600 0.906 4.338

0.800 0.793 4.957

1.000 0.634 6.197 (S1)(Fv) = SD1

1.200 0.529 7.436

1.400 0.453 8.675

1.600 0.396 9.915

1.800 0.352 11.154

2.000 0.317 12.393

2.200 0.288 13.633

2.400 0.264 14.872

2.600 0.244 16.111

2.800 0.227 17.350

3.000 0.211 18.590

Based on the long acceleration coefficient SD1 value of 0.634 g calculated as Fv.S1, the bridge site is assigned to seismic hazard Zone 4 according to Table 3.10.6-1- Seismic Zones in AASHTO. Seismic hazard zones reflect the variation in seismic risk in different regions needing different requirements for design as depicted in Table 4.7.4.3.1-1 in AASHTO.

0.2

0.4

0.6

0.8

1.2

0 0.5 1 1.5 2 2.5 3

S pe ct ra l A cc el er at io n, S a (g

Period, T (sec)

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The uniform hazard spectrum presented above is most commonly used for seismic design of ordinary structures and is developed based on representative earthquake magnitudes for the region. In high seismic zone areas such as this project, a detailed evaluation of individual magnitude and distance of earthquakes most contributing to the seismic hazard spectrum known as "deaggregation" may be performed for each spectral period on the uniform hazard spectrum.

Deaggregation was performed for this project to determine the nearest source event.

The magnitude associated with the design ground motion is required for site response or time domain structural analysis. The 2008 USGS Interactive Deaggregation for the Conterminous United States was used to establish the design magnitude based on the peak ground acceleration. For consistency with AASHTO, the deaggregation was conducted for a ground motion with a 5% probability of occurrence in the 50 year period, a ground motion with a return period of approximately 975 years. This return period is consistent with the 1000 year return period for the AASHTO seismic hazard map.

A histogram for the deaggregation hazard is shown in Illustration 2. The histogram shows that 97% of the hazard is from earthquakes of magnitude 5.8 to 7.4 in subduction zones and 27% of the hazard is from a magnitude 6.4 to 8.0 on individual faults 3 to 24 miles from the site. The most prominent individual fault near the bridge site is the San Andreas fault, which is located approximately 20 miles east of the site. The California Geologic Survey (2008) classifies faults into “A”, “B”, or “C” faults or fault zones, decreasing in potential earthquake magnitude and impact. The San Andreas fault is considered a “A” fault by the California Geologic Survey (2008) and described as a dextral (right-lateral) strike-slip fault with a maximum earthquake magnitude of 8.01 and an annual slip rate of greater than 5.0 mm/year. Based on these data, the representative design earthquake was established as a magnitude 7.99 event 20 miles from the site along the subduction zone off the coastline.

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ILLUSTRATION 2:- Histogram for the Deaggregation Hazard Illustrating the Potential Source Distances from the Bridge Site

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SECTION THREE--SUBSURFACE INVESTIGATION

3.1 FIELD EXPLORATION

A field exploration program consisting of two borings at the proposed bridge abutment locations was conducted by CFL between June 24 and June 26, 2014. Drilling operations consisted of hollow stem auger and mud rotary methods and conducting Standard Penetration Tests (SPT’s) at 5-foot intervals. Borings were logged in the field by the engineering geologist of CFL, and soil samples were classified in the field and transported to the CFL Materials Laboratory in Lakewood Colorado for testing. Details of the field exploration program and the boring logs are included in Appendix A. A summary of the boring locations and depths are shown in Table 4 Summary of Bridge Borings.

TABLE 4: Summary of Bridge Borings

Borehole Number

Location Station and Offset (ft) Ground

Elevation (ft) Boring

Depth (ft)

B-1 Abutment 1

(West Abutment) STA 1+92 8-ft LT 226.20 73.9

B-2 Abutment 2

(East Abutment) STA 3+33 10-ft LT 223.50 73.9

3.2 LABORATORY TESTING

Laboratory testing was conducted on select soil samples obtained during the subsurface investigation to determine material properties for use in the foundation design. Laboratory test procedures and results are presented in Appendix B, Laboratory Test Results and summarized in Table 5, Summary of Soil Classification Index Tests. Samples showing similar material properties were combined for a more representative testing size.

TABLE 5: Summary of Soil Classification Index Tests

Boring Station and Location

SPT Sample Sample Depth (feet)

Percent Passing No. 200

LL PI USCS AASHTO

B-1 1+92 West

Abutment

SPT-1, SPT-2 5.0-11.5 7 NV NP SW-SM A-1-b(0) SPT-3, SPT-4 15-21.5 4 NV NP SW A-1-b(0)

SPT-5 25-26.5 4 NV NP SW A-1-b(0) SPT-6 through SPT-8 30-38.5 22 NA NA NA NA

SPT-9, SPT 10 43-48.8 44 NV NP SM A-4(0)

SPT 11 through SPT 15 53-73.9 53 26 2 ML A-4(0)

B-2 3+33 East

Abutment

SPT-1, SPT-2, 5.0-11.5 22 NV NP SM A-2-4(0)

SPT-3 15-16.5 8 NA NA NA NA

SPT-4 through SPT-6 20-31.5 26 NV NP SM A-2-4(0)

SPT-7 35 -36.5 80 43 26 CL A-7-6(20)

SPT-8 through SPT-11 40-54 56 NV NP ML A-4(0) SPT-12 through SPT-15 60-74 64 30 8 CL A-4(3)

NV = No Value, NP = Non-Plastic, NA = Not Applicable/no tests conducted

Electrochemical tests were conducted on one representative sample obtained from both B-1 and B-2 collected from the site to determine soil corrosivity and potential detrimental effects on

9 | P a g e concrete and buried metals. Testing for resistivity and pH were performed in general accordance with AASHTO T 288 and T 289, respectively. Tests for sulfate and chloride content were also performed due to resistivity results less than 5000 ohm-centimeters in accordance with AASHTO T 290 and T 291, respectively. A discussion of corrosion requirements, electrochemical test results, and recommendations for corrosion restrictions is presented in Section 4.3, Corrosion Potential. A summary of analytical test results are provided in Table 6, Summary of Electrochemical Properties.

TABLE 6: Summary of Electrochemical Properties

Boring Station Sample

Depth (ft)

Class.

(AASHTO)

(USCS)

pH

Resistivity (ohm-cm)

Sulfates (%/ppm)

Chlorides (%/ppm)

B-1 &B-2 varies 30-74 NA 6.0 1,660 0.074/740 0.0026/26 NA = Not Applicable/no test conducted

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SECTION FOUR--ANALYSIS AND RECOMMENDATIONS

For the analysis and foundation design of the proposed bridge, a geotechnical profile was developed from surface observations and laboratory testing conducted during the geotechnical investigation. Bridge foundation design, corrosion potential, shrink/swell recommendations, and construction considerations are discussed in the following sections. All foundations are in accordance with the American Association of State Highway and Transportation Officials (AASHTO) Load and Resistance Factor Design (LRFD) Bridge Design Specifications, 6th Edition, 2012 with 2013 interim revision.

4.1 BRIDGE FOUNDATION DESIGN

The proposed new structure is approximately 100-feet long simple-span bridge with prestressed box beam girders and cast-in-place concrete deck that will be both economical and also a fast construction schedule. Approach slabs will not be constructed at this site.

Since alignment and width of the proposed bridge will be constructed to match the existing roadway template, no retaining structures are required. The proposed bridge width is measured normal to the construction layout line and constant across the length of the bridge. The proposed bridge deck width includes two 8-foot, 1-inch lanes, with no proposed shoulders.

The replacement bridge will be located at a slightly higher elevation as the existing embankment fills at existing abutment locations. The proposed bridge deck elevation at centerline is 225.96 feet at Abutment 1 and 224.15 feet at Abutment 2. The channel invert elevation for El Toro Creek at the centerline of the proposed bridge is approximately 214.2 feet. Proposed abutment centerline stations and bottom of abutment and pile cap elevations are shown in Table 7, Proposed Bridge Foundation Locations.

TABLE 7: Proposed Bridge Foundation Locations

Foundation Approximate

Station, ft Top of Pile Cap

Elevation, ft Bottom of Pile Cap Elevation, ft Abutment 1-West STA 2+10.00 222.2 219.2 Abutment 2-East STA 3+10.00 220.5 217.5

4.1.1 Geotechnical Profile

In general, three distinctive soil layers were encountered in each of the abutment borings at similar elevations at both abutments. Detailed descriptions of the bridge borings and boring logs are included in Appendix A. These materials were generally characterized for conservative design purposes as the following:

1. Layer 1: Brown to dark brown sand, some gravel and silt classified as SM and SW-SM.

2. Layer 2: Brown clay with sand classified as CL-ML;

3. Layer 3: Brown hard to very hard sandy silt and sandy clay layers classified as cohesive intermediate geomaterials (fine grained sedimentary hard soils), IGM

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Generally, the subsurface materials encountered were a loose non-plastic medium to fine grained sands layer with varying percentages of silts and gravels existed below the gravel surfacing to approximately 32 feet. An approximately 7-foot thick medium dense highly plastic clay layer with sands was encountered in one of the borings between 31.5 feet and 38 feet. Hard soils/soft rock classified as IGM (intermediate geomaterials) with SPT consistently measuring greater than 50 blows per foot were encountered below the sandy clay layer to approximately 74 feet below the ground surface in both of the borings.

The sandy soils are generally classified as SM, SW-SM, and SW in accordance with the Unified Soil Classification System and as A-1-b and A-2-4 in accordance with the AASHTO soil classification system. The clayey sandy soils are generally classified as CL and ML in accordance with the Unified Soil Classification System and as A-7-6 in accordance with the AASHTO soil classification system. The SPT-N values were corrected for an assumed hammer efficiency (80 percent) and overburden pressure to obtain the N1(60) values for LRFD foundation design. A summary of the steps used for determining a representative N1(60) value for deep foundation design purposes are shown in Table 8.

SPT N-values (corrected) within the sandy soils range from 13 to 41 with the higher values occurring when gravels were encountered. SPT N-values were corrected for an automatic hammer efficiency of 80 percent (assumed) and the overburden pressure. SPT N-values (corrected) within the clay/silt soils range from 18 to 66. SPT’s are generally less reliable for predicting strength and compressibility for clays. IGM formations could be either soil that has densified or strengthened or intrinsically weak rock that have undergone a limited amount of gravitational compaction and cementation. The IGM is penetrable with the SPT and should be expected to be good foundation materials for piles.

The groundwater table was encountered at an approximate elevation of 210.5 to 213.2 feet at the time of drilling, roughly halfway through the sandy materials; therefore, pore-water pressure is partially included in calculating the effective overburden stress (below 13 feet depth).

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TABLE 8: N1(60) Calculation for Overburden Material at Bridge Location

Boring SPT Depth1, ft σv' 2,ksf N3 N60

Cn

5 N1(60)

N1(60)

Less Refusal

B-1

1 5 600 7 9.3 1.40 13.1 13 2 10 1200 12 16.0 1.17 18.8 18 3 15 864 10 13.3 1.28 17.1 17 4 20 1152 26 34.7 1.19 41.1 Neglect 5 25 1440 11 14.7 1.11 16.3 16 6 30 1728 Refusal Refusal 1.05 Refusal Neglect 7 33 1736 Refusal Refusal 1.05 Refusal Neglect 8 38 1999 Refusal Refusal 1.00 Refusal Neglect 9 43 2907 Refusal Refusal 0.88 Refusal Neglect

10 48 3245 Refusal Refusal 0.84 Refusal Neglect 11 53 3583 Refusal Refusal 0.81 Refusal Neglect 12 58 3853 Refusal Refusal 0.78 Refusal Neglect 13 63 4259 Refusal Refusal 0.75 Refusal Neglect 14 68 4597 Refusal Refusal 0.72 Refusal Neglect 15 73 4935 Refusal Refusal 0.70 Refusal Neglect

B-2

1 5 600 15 17.3 1.40 24.3 24 2 10 1200 11 14.7 1.17 17.2 17 3 15 864 9 12.0 1.28 15.4 15 4 20 1152 24 32.0 1.19 38.0 Neglect 5 25 1440 10 13.3 1.11 14.8 14 6 30 1728 11 14.7 1.05 15.4 15 7 33 1736 21 28.0 1.05 29.4 29 8 38 1999 14 18.7 1.00 18.7 18 9 43 2907 Refusal Refusal 0.88 Refusal Neglect

10 48 3245 Refusal Refusal 0.84 Refusal Neglect 11 53 3583 Refusal Refusal 0.81 Refusal Neglect 12 58 3853 Refusal Refusal 0.78 Refusal Neglect 13 63 4259 Refusal Refusal 0.75 Refusal Neglect 14 68 4597 Refusal Refusal 0.72 Refusal Neglect 15 73 4935 Refusal Refusal 0.70 Refusal Neglect

1Begining depth of SPT-N reading 2Effective overburden pressure 3Field SPT-N value 4Calculated using AASHTO equation 10.4.6.2.4-2

5Calculated using AASHTO equation 10.4.6.2.4-1 6Calculated using AASHTO equation 10.4.6.2.4-3 7Refusal values neglected in determining average N1(60) The spt reached 50 blows prior to penetrating 6 inches – these values were neglected.

Although cobbles and boulders were not encountered in the borings, they should be anticipated as they are present on the surface in the active stream channel. Material properties estimated for foundation design were based on the field investigation results, laboratory test results, and presumptive empirical strength characteristics of similar materials. Selected design values are presented in Table 9, Estimated Subsurface Material Properties for Bridge Foundation Design.

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TABLE 9: Estimated Subsurface Material Properties for Bridge Foundation Design

Depth Below Ground Surface (feet)

Material Description Total Unit

Weight ɣ, (pcf)

Effective Unit

Weight* ɣ’, (pcf)

Cohesion c

(psf)

Friction Angle Փ

(deg)

0-32.0 SAND 120 57.6 0 32

32.0-39.0 CLAY 110 47.6 100 23

39.0 to 74.0 IGM 130 67.6* 200 38

Notes: *Groundwater estimated at 13 feet. Effective unit weight below the estimated groundwater surface was used in the calculations.

For IGM materials, effective unit weight was discounted due to density.

The modulus of elasticity for normally consolidated granular soils tends to increase with depth.

The modulus of elasticity for soil (Es) can be assumed to increase linearly with depth starting at zero at the ground surface. The rate of increase of soil modulus is applied per foot of overburden

(z) as shown in the equation Es=0.556z, per AASHTO 10.4.6.3.

4.1.2 Groundwater Conditions

Groundwater elevations are generally required for evaluating side and tip resistance under axial loads, computing lateral loads, and determining driven pile construction procedures in cohesionless soils. Groundwater was encountered during drilling close to the elevation of the stream channel. Groundwater depth at the time of drilling was measured at about 13 feet below the ground surface), as indicated on the boring logs in Appendix A. Fluctuations in the groundwater level due to seasonal and climatic effects was not monitored, but should be expected. Based on the subsurface investigation and site observations, groundwater should be anticipated during foundation construction and pile driving operations. For design purposes, groundwater was assumed to be at the stream elevation which corresponds.

4.1.3 Liquefaction Potential

The potential for soil liquefaction in cohesionless soil deposit may exist when the soil loses a substantial amount of strength due to pore pressure generation resulting from cyclical earthquake. Loose, cohesionless soils tend to compact during earthquake cyclical shaking and this tendency causes the pore water pressures in the solid to increase until the pore water dissipates from the soil layers. This in turn, causes a reduction in effective stress and associated reductions in soil strength and stiffness. Submerged, recently deposited, shallow, loose sand and silty sands are particularly susceptible to liquefaction. Dense natural, well-compacted fills and cohesive soils are known to have a relatively low susceptibility to liquefaction.

With a site peak ground acceleration (As) of 0.427g in Seismic Zone 4, a water table close to the ground surface, and loose sandy soils with low SPT-N values, the site can be characterized as potentially liquefiable for Layer 1. An evaluation for the potential for, and consequences of, liquefaction in soils near the surface were made in accordance with the criteria (FHWA, 2011).

The guidelines indicate that sandy soils (Layer 1) are considered liquefiable if the minimum corrected Standard Penetration Test resistance (N1)60 values are less than 30 blows per foot, the layer is submerged and is less than 75-feet deep.

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Soil liquefaction analysis indicates that the shallow, submerged, loose, sandy soil SM-SW layer met the SPT (N160 average values < 30), depth (< 75 feet), and groundwater elevation (submerged) criteria for potential liquefaction. Therefore the sandy layer (Layer 1) between the ground surface and 32-feet is anticipated to liquefy during an extreme event and should not be considered for skin friction at the Extreme Event I Limit State.

4.1.4 Liquefaction Induced Settlement

A consequence of earthquake induced liquefaction is the subsequent post-earthquake dissipation of excess pore-water pressure, potentially resulting in soil settlement and downdrag loads. The top sand layer (0 to 32 feet) is considered liquefiable (N160 average values is 16), depth (32 feet), and groundwater elevation (13 feet) and was further analyzed for post-liquefaction and induced settlement. Based on experimental studies, Tokimatsu and Seed (1987) developed a simple chart based procedure for estimating post liquefaction settlement. The Tokimatsu and Seed (1987) chart for a magnitude 7.5 earthquake plots the volumetric strains corresponding to the blow counts (N160) and cyclic stress ratios (CSR) for each liquefied layer should be multiplied by the layer thickness to estimate settlement. Following this method (as recommended in FHWA, 2011), maximum settlement for the sandy layer if liquefied, was estimated at approximately 7.8 inches (attached in Appendix E).

4.1.5 Pre- and Post-Liquefaction Properties for Lateral Resistance

Lateral pile resistance for static and seismic loading at the Service and Extreme Event I Limit States should be evaluated by the structural engineer using the parameters in Table 9. These material strength values were estimated based on recommendations published by Seed and Harder (NCHRP 12-70). These investigators found little evidence of cyclic strength loss in cohesive soils during earthquakes and noted that cohesive soils also tend to show an increase in undrained shear strength when subject to rapid cyclical loading. They concluded that for earthquakes of greater than magnitude 7.0, a strength loss of 10-15 percent is recommended.

Also Makdisi and Seed (1978) found that substantial permanent strains may be produced by cyclic loading of clay soils to stresses near the yield stress, while essentially elastic behavior is observed for large numbers of cycles of loading at cyclic shear stresses of up to 80 percent of the untrained strength. These researchers recommend the use of 80 percent of the undrained strength as the “dynamic yield strength” for soils that exhibit small increases in pore pressure during cyclic loading, such as clayey materials. Typically, based on tests by Seed and Others, saturated sands that liquefy possess some “residual” shear strength even when in the liquefied state. In initially loose soils, the residual strength may be very small and of little consequence. The seismic (post-liquefaction) parameters in Table 10 assume a 20 percent reduction in strength and stiffness for the liquefiable sand layer (Layer 1).

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TABLE 10:- Soil Characterization for Pre- and Post-Liquefaction

LAYER 1

SAND

LAYER 2

CLAY

LAYER 3

IGM

Top Elevation 223ft 192 ft 185 Bottom Elevation 192 ft 185 ft 149

Material Description (1) Sand Clay IGM Layer Thickness 32 ft 7 ft 35 ft

STATIC (PRE-LIQUIFACTION) PARAMETERS

Undrained Strength, su NA 2,000 psf NA Friction angle (deg.) 32 23 38

Unit weight, (2) /’ (pcf) 120/57.6 pcf 110/47.6 pcf 130/67.6 pcf

LPile PLUS 3.0 soil type Medium Submerged

Sand Stiff Clay

Sand (Soft Rock/IGM)

Coef. of subgrade reaction, k, 60 pci 500 pci 900 pci Strain @ 50%, 50 NA 0.007 NA Strain @ 100%, 100 NA 0.01 N/A

SEISMIC (POST-LIQUEFACTION) PARAMETERS

Undrained strength, su NA 2,000(3) psf N/A

Coef. of subgrade reaction, k, 48(4) pci 500(3) pci NA Strain @ 50%, 50 NA 0.007 NA Strain @ 100%, 100 NA 0.01 N/A

1. Review boring logs for further information on material types.

2. Water is at approximately at 210.5 ft elevation at the abutments.

3. Based on no reduction in strength following strong seismic shaking.

4. Assume 20% reduction in strength in layer 1 following liquefaction.

Lateral spreading following an earthquake may occur beneath a bridge approach or abutment fills if the under laying soil is considered liquefiable. The static factor of safety of the soil mass near the bridge abutment is greater than one but when lateral movements caused by an earthquake induce shear stresses that exceed the soil shear strength of the liquefied soil, lateral spreading may occur.

4.1.6 Bridge Foundation Selection

Driven (non-displacement) steel H-piles were selected for the foundation of this bridge after several options were considered. A shallow foundation system, including Geosynthetic Reinforced Soil - Integrated Bridge System (GRS-IBS) was considered and determined to be unfeasible due to the bridge location, bridge length, and depth of design scour and liquefiable soil layer. Estimated scour depth for a Q50 bridge design criteria are; the design scour (100 year) elevation is 207.0 feet and the check scour (500 year) elevation is 198.0 feet.

Deep foundation options were evaluated based on feasibility and costs. Driven piles, drilled shafts, and micropiles were all considered feasible at this project site based on site subsurface materials encountered consisting of sandy materials with silts and gravels. Based on local construction practices and lowest foundation costs, driven steel H-piles are the preferred alternative for the foundation at the bridge abutments. Despite the possibility of encountering cobbles and boulders in the overburden material, it is anticipated that pile driving shoes and heavier piles than required for design will allow the H-piles to be driven to depths necessary to satisfy loading and scour requirements. Drilled shafts and micropiles were considered as deep

16 | P a g e foundation alternatives to driven piles, but the cost of both options is higher than driven piles.

Driven piles are capable of supporting heavy concentrated loads due to their ability to resist axial and lateral loads. Piles are designed to have adequate geotechnical and structural resistances with a tolerable settlement.

4.1.7 LRFD Driven Pile Foundation Design

Driven steel H-piles size 14x89 was the analyzed as the main foundation system for the bridge abutments. The heavier pile was selected due to the potential for corrosion loss over the service life of the pile, for driving through gravels and cobbles, for better lateral resistance in liquefiable sand, and for greater skin friction development at shallower depths. Pile design load is supported by soil side resistance developed only in soil layers that contribute to long-term load support.

Soil resistance from materials subject to scour and liquefaction are not considered for the axial resistance calculations.

Based on the AASHTO LRFD (6th edition) approach provided nominal axial loads should be factored by the appropriate limit state resistance factor (φ) for strength, service, and extreme event. Factored resistances are determined from estimated nominal pile resistances and resistance factors that are based on confidence in soil parameters, construction methods, pile design methods, and level of construction monitoring. The recommended resistance factor for axial compression at the Strength I Limit State is 0.65 when dynamic testing and signal matching are performed. If no dynamic testing is performed, then the recommended resistance factor for axial compression at the Strength I Limit State is 0.4. The resistance factors for the Service I and Extreme Event I and II Limit States are 1.0. These factored loads must not exceed the factored resistances at all limit states for stability. Because soil Layer 1 is susceptible to strength loss during strong ground shaking or liquefaction, the Extreme Event I Limit State should be evaluated for compressional loading and should account for a downdrag load within the top layer which is produced due to material strength loss caused by temporary pore-water pressure build up in the soil.

Four cases were evaluated to determine the controlling loading case for the design of the driven piles. Two cases in the Strength I limit state were considered, one with the heaviest loading and the other with the 100-year scour considerations. Two Extreme Event cases were evaluated, one in Extreme Event I limit state considering liquefiable materials, and the other in the Extreme Event II limit state considering the 500-year check scour event. Each case is presented in Table 11, Summary of Nominal Pile Resistance per Pile below.

4.1.8 Downdrag Load

To mitigate the potential for post-liquefaction settlement of the bridge at the Extreme Event I Limit State, the bearing resistance developed below the liquefaction (32 feet) should be greater than the sum of the design and the downdrag loads. The downdrag load was calculated as 70 kips based on the potential strength loss in the liquefiable layer 1 following an earthquake. This should be treated as an extreme event loading state and no further reduction in resistance should be applied (resistance factor of 1.0). In accordance to AASHTO guidelines a load factor of 1.05

17 | P a g e should be applied to the downdrag loads. If this available geotechnical resistance is exceeded, and the structural pile resistance is not, settlement of the pile should be expected following liquefaction. If the structural resistance is exceeded by the combination of downdrag and structure loads, lasting pile damage can be expected. The nominal pile resistance available to support the structure plus downdrag loads is estimated by considering the positive skin friction and tip resistance below the layer contributing to downdrag.

To mitigate the liquefaction effects on stability, downdrag loads can be reduced by either isolating the upper part of the pile from the soil with low friction material or a pipe, or the geotechnical resistance can be increased by driving the piles deeper into the IGM. The results indicate nominal friction resistance develops at a rate of approximately 30 kips per foot for the 14x89 pile in IGM. Each loading condition case is listed below in Table 11, Summary of Nominal Pile Resistance per Pile.

TABLE 11: Summary of Nominal Pile Resistance (Target Resistance) per Pile

Case Loading

Condition

Nominal Pile Driving Resistance (Target Resistance) (kips/pile)

Required Depth (Elevation) for Maximum Factored Load (feet) Estimated

Settlement (inches)

PDA

= 0.65

PDA

= 1.0

No Testing = 0.4

PDA

= 0.65

PDA

= 1.0

No Testing = 0.4

Strength I Limit State No Scour Large Vehicle

– Heaviest Load

231 - 375** 36.1

(187.9)

49.2 (174.8)

<0.1

Strength I Limit State Design Scour (100 Years)

169 - 275 35.2

(188.8)

43.7 (180.3)

<0.1

Extreme Event II Limit State Check Scour (500 Years)

- 100 NA - 34.0 NA <0.1

Extreme Event I Limit State Seismic Event - Liquefaction

- 228 + 70* NA - 36.3 NA <0.1

Notes * Downdrag (70 K/pile) is added to Case 4 after liquefaction.

** Controlling design case

It is assumed no pile driving testing will be performed due to cost savings in construction since lateral load capacity of piles is the governing component which requires a minimum tip elevation of 172 feet. The resistance factor for the Strength Limit State I axial compressional design will be 0.4 in accordance with the AASHTO Table 10.5.5.2.3-1. The analysis results for all cases are presented in Appendix D. For consistency, all plots were plotted from the ground elevation during drilling, not the top of the pile cap.

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4.1.9 Nominal Axial Resistance

The static nominal axial resistance of a single pile in compression at each abutment was calculated using the APILE Plus computer program. In the APILE program, the soil profile consisting of the soil unit weights and strength parameters including the percentage strength loss during driving is input. For the selected pile type, the program calculates the pile resistance versus depth of the entire soil profile using three methods: the American Petroleum Institute (API), the FHWA method, and the US Army Corps of Engineers (ACE) method. The results from all three analysis methods were plotted and are presented in the appendices. Generally, the FHWA method for cohesionless soils using the Nordlund/Thurman Method for frictional resistance and end bearing resistances based on blow counts and estimated friction of the soil provides the most capacity per unit length and uses the same effective stress equations as the old DRIVEN program. The proposed 14x89 steel H-pile section was analyzed using the FHWA method with the resistance gained solely through frictional side resistance (skin friction) using the geotechnical profile material properties presented in Table 9 and blow counts from the boring logs presented in Appendix A. The box perimeter and box area were used for pile perimeter and tip area calculations, respectively.

Due to potential channel degradation scour at the bridge site, the reduction in pile capacity from loss of pile resistance in the scour zone and the influence of the reduced effective overburden pressure from soil removal on the resistance calculated in the underlying layers was considered in the APILE analysis. Layer 1 is assumed liquefiable and the loss of material strength was assumed in the calculations in Case 4 for Extreme Event I state.

Geotechnical losses were not considered when calculating the required driving resistance due to cohesionless properties of the on-site overburden material, as recommended in the FHWA Design and Construction of Driven Pile Foundations reference manual (2006). Geotechnical losses represent the difference in resistances between the nominal and driving conditions at depth and generally impact driven pile design in cohesive soils below the groundwater table.

Additionally, downdrag forces were considered due to the seismic effects on soil liquefaction. It is anticipated that all embankment fills are constructed and compacted to plan elevation prior to pile driving. Previous Table 11 provides a summary of the APILE calculated nominal pile resistance per pile for each loading case and Table 12, Summary of Pile Recommendations, summarizes pile design recommendations based on the APILE analysis.

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TABLE 12: Summary of Pile Recommendations

Recommendations Abutments 1 and 2

Pile Resistance Verification No Testing

Resistance Factor for Driven Piles 0.4

Pile Type 14x89 H-Pile Pile Material Minimum of 50 ksi Steel

Number of Piles per Abutment 4

Compressive Nominal Bearing Resistance Required During Driving (Includes Side Resistance Below Design Scour Elevation)

375 kips per pile (based on Case 1, Table 11, Strength I Limit State/ No Scour w/ heaviest load)

Bottom of Pile Cap Elevation 219.23 feet @ Abut.1 & 217.55 @ Abut.2

Estimated Length of Piling to Achieve Required Geoetchnical Nominal Bearing Resistance

(including embedment of 1 foot into each pile cap)

49.2 feet

Estimated Pile Tip Elevation Required to Achieve Required Geoetechnical Nominal Bearing Resistance

(not accounting for bending in the pile due to liquefaction case)

174.8 feet

Minimum Pile Tip Elevation (accounting for bending in the pile from bridge LPILE analysis due to liquefaction case) 172 feet

4.1.10 Pile Setup

Due to the subsurface materials encountered, setup will not likely make a major contribution to the long-term resistance in the sand and clay materials and should not be accounted for in determining nominal resistance for the pile foundation at the Guidotti Bridge site.

Pile setup is an increase in the nominal axial resistance that develops over time, predominantly along the pile length. When soils are compressed and disturbed due to pile driving, excess pore pressures develop. These excess pore pressures are generated partly from the shearing and remolding of the soil and partly from radial compression as the pile displaces the soil. The excess pore pressures cause a reduction in the effective stresses acting on the pile, and thus a reduction in the soil shear strength. This results in a reduced pile capacity during, and for a period of time after, driving. After driving, the excess pore pressures will dissipate primarily through radial flow of the pore water away from the pile. With the dissipation of pore pressures, the soil reconsolidates and increases in shear strength. This increase in soil shear strength results in an increase in the static pile resistance. In sand and gravel, pore pressure generally dissipates rapidly and restrike is not anticipated to significantly improve capacity.

4.1.11 Group Effects on Axial Resistance

The resistance of a pile group to the applied axial loads is not necessarily the sum of the axial resistance of individual piles within the group. The zone of influence from an individual pile in a pile group may intersect with other piles, depending on the pile spacing. Historically the efficiency of groups of piles has not been a concern as long as the center-to-center spacing between piles is greater than 2.5 times the pile diameter (2.5B) or 30.0 inches to avoid interference between adjacent piles during construction. Therefore, an efficiency factor (η) is 1.0

20 | P a g e for the pile group with center-to-center spacing of 2.5 diameters or more. It is anticipated that there will be five piles at each abutment spaced approximately 5 feet apart, which corresponds to an efficiency factor of 1.0.

4.1.12 Lateral Loads

Lateral soil-structure interaction analysis of the driven piles is performed by CFLHD bridge engineers on single and group piles. For lateral loading analyses use the subsurface soil parameters provided in Table 9 of this report to estimate the lateral load displacement behavior.

The values presented in this subsection do not include reduction factors or load factors. A strength limit state resistance factor of 1.0 is specified in AASHTO (2012) for lateral geotechnical resistance of a single pile or pile group. Factored strength limit state lateral loads should be used in the analysis. A lateral load analysis should be conducted on a single pile from each abutment using the table soil parameters for submerged “medium sand”.

The soil behavior around the laterally loaded pile is described by the soil P-y curves. Material properties in Table 9 are for single piles and do not account for the reduced lateral resistance of piles in a group. P-multipliers are a function of the number of rows of piles and center-to-center pile spacing in the direction of loading. P-multipliers are required even for a single row of piles if the center-to-center spacing of the piles is less than 5 pile diameters. P-multipliers are specified in Table 10-7.2.4-1 in AASHTO (2012). When this analysis method is used, the resistances at the strength limit state as represented by the P-y curves should not be factored since they already represent the nominal conditions.

4.1.13 Pile Driveability

To assure pile drivability without damage, the pile must satisfy two aspects of drivability: (1) the pile must have sufficient stiffness to transmit driving forces large enough to overcome soil resistance, and (2) the pile must have sufficient structural strength to withstand the driving forces without damage due to overstressing. A pile with a high material strength can tolerate higher driving stresses, which may allow for the use of a larger pile hammer. This may allow a slightly higher pile resistance to be obtained before refusal driving conditions or pile damage occurs.

Even if the pile structural capacity and geotechnical capacity both indicate a high pile resistance could be used, a high pile resistance may still not be obtainable because driving stresses may exceed allowable driving stress limits.