Final_Geotechnical_Report_Don_Edwards__rev_ision_Final_January2015.pdf

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LA RIVIERE MARSH BRIDGE

DON EDWARDS SAN FRANCISCO BAY

NATIONAL WILDLIFE REFUGE

CA FLAP SR 84(U)

Final Geotechnical Design Report Report # CA-RX-0014-01

CA Bridge Number 33-0757Y

Geotechnical Services Branch August 2014 i

Table of Contents

SECTION ONE - INTRODUCTION

1.1 BACKGROUND AND LOCATION

1.2 SCOPE AND PURPOSE

SECTION TWO – GEOLOGY AND SEISMICITY

2.1 REGIONAL GEOLOGY

2.2 SEISMIC DESIGN PARAMETERS

SECTION THREE–SUBSURFACE INVESTIGATION

3.1 FIELD EXPLORATION

3.2 SUMMARY OF SUBSURFACE EXPLORATION

SECTION FOUR – ANALYSIS AND RECOMMENDATIONS

4.1 SITE CHARACTERIZATION

4.1.1 GEOTECHNICAL PROFILE

4.1.2 SCOUR

4.2 LIQUEFACTION POTENTIAL

4.2.1 LIQUEFACTION POTENTIAL

4.2.2 LIQUEFACTION INDUCED SETTLEMENT

4.2.3 PRE- AND POST-LIQUEFACTION PROPERTIES FOR AXIAL AND LATERAL RESISTANCE

4.2.4 LATERAL SPREAD DISPLACEMENT

4.3 BRIDGE FOUNDATION DESIGN RECOMMENDATIONS

4.3.1 NOMINAL COMPRESSION RESISTANCE

4.3.2 NOMINAL LATERAL PILE RESISTANCE

4.3.3 GROUP EFFECTS

4.3.4 SETTLEMENT

4.3.5 PILE DYNAMIC TESTING

4.3.6 PILE DRIVABILITY

4.4 ABUTMENT DESIGN

4.5 CONSTRUCTION CONSIDERATIONS

4.6 DISCLAIMER/LIMITATIONS

SECTION FIVE -REFERENCES

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TABLES

TABLE 1 Summary of Seismic Parameters Corrected for Site Class D

TABLE 2 Summary of Design Response Spectrum Data for Site Class D

TABLE 3 Summary of Laboratory Results from Borings at La Riviere Marsh Bridge

TABLE 4 Soil Characterization

TABLE 5 Initial Screening of Liquefaction Potential

TABLE 6 Foundation Recommendations for Each Abutment and Pier Pile

TABLE 7 Pile P-Multipliers, Pm for Multiple Row Shading

FIGURES

FIGURE A Design Response Spectrum Corrected for Site Class D

FIGURE 1 Idealized Soil Profiles at Piers and Abutments

FIGURE 2 Nominal Bearing Resistance for the Piers Using Driven, 356-mm concrete pile

FIGURE 3 Nominal Bearing Resistance for the Abutments Using Driven, 356-mm concrete pile

FIGURE 4 Nominal Bearing Resistance for the Piers Using Driven, 406-mm concrete pile

FIGURE 5 Nominal Bearing Resistance for the Abutments Using Driven, 406-mm concrete pile

FIGURE 6 Nominal Bearing Resistance for the Piers Using Driven, 457-mm concrete pile

FIGURE 7 Nominal Bearing Resistance for the Abutments Using Driven, 457-mm concrete pile

FIGURE 8 Comparison between the Nominal Resistances estimated from CPT data and Driven for Abutments, 356-mm concrete pile

FIGURE 9 Comparison between the Nominal Resistances estimated from CPT data and Driven for Piers, 356-mm concrete pile

APPENDICES

APPENDIX A Project Location Maps

APPENDIX B Borehole Logs and Photos

APPENDIX C Cone Penetrometer Data

APPENDIX D Laboratory Data

APPENDIX E Analysis

APPENDIX F Special Contract Requirements

La Riviere Marsh Bridge, Don Edwards San Francisco Bay NWR Project, CA FLAP SR 84(U)

SECTION ONE - Introduction

1.1 BACKGROUND AND LOCATION

The Don Edwards San Francisco Bay National Wildlife Refuge was founded to protect and preserve the wildlife habitat and the threatened and endangered species that exist in the area. Salt marshes, salt ponds, and mudflats surround the Refuge, and tidal sloughs which attract wildlife indicative to the area.

Migratory birds that use the area for nesting and feeding largely inhabit the Refuge. The Refuge provides a variety of interests through their Environmental Education Center and multitude of trails for hiking, viewing, fishing and boating.

The Refuge is located southern part of the San Francisco Bay primarily south of State Highway (SH) 84, north of the Bayshore Freeway and SH 237, and east of US 880. The boundary of the Refuge varies considerably within this general area. The main entrance to the Refuge is from the north (near Fremont) called Marshlands Road. Marshlands Road extends west from Thornton Avenue (entrance to the

Refuge) to the Dumbarton Fishing Pier near the edge of San Francisco Bay. Except during nesting season, Marshlands Road is frequently used by anglers for fishing at the pier and by locals who use the road as their daily bicycle-commuting route. La Riviere Marsh Bridge is located on Marshlands Road in the Refuge.

The rehabilitation work for the Refuge includes resurfacing about one kilometer of Marshlands Road from Thornton Avenue to the entrance to the Environmental Education. The current roadway width varies due to existing medians and pedestrian lane along the route. Currently, the roadway consists of two-3.7-meter lanes, with a 1.2-meter pedestrian lane on each side. The pedestrian lane extends along the entire project length, except on the La Riviere Marsh Bridge. The existing, 27-meter long bridge is located approximately 350-meters west of the Refuge entrance and are approximately 7-meters wide from curb to curb (or 7.9 meters from rail to rail). The proposed rehabilitation includes a replacement bridge to accommodate a pedestrian lane.

In January and March 2003, the Geotechnical Section of the Central Federal Lands Highway Division

(CFL) of the Federal Highway Administration (FHWA) conducted a subsurface foundation investigation for the proposed bridge replacement at La Riviere Marsh. A location map is included in Appendix A.

1.2 SCOPE AND PURPOSE

The scope of work includes subsurface investigation and analysis for determining the subsurface soil profiles and ground water conditions at each abutment and bridge pier location, determining material physical properties of representative subsurface samples for static and seismic foundation design, and providing AASHTO LRFD based design recommendations on foundation types/geometry, bearing resistance, settlement, global stability, and earth pressures. This report presents the findings of the subsurface investigation programs and provides geotechnical recommendations for the design and construction of the La Riviere Marsh Bridge. Specifically, the purposes of this geotechnical investigation are as follows:

Characterize the site and subsurface conditions at the proposed pier and abutment locations for the Bridge.

Identify soil parameters for foundation design at the Strength I, Service I and Extreme Event I

Limit States.

Make foundation design recommendations consistent with AASHTO LRFD and local design guidelines.

Provide recommended Special Contract Requirements.

SECTION TWO – Geology and Seismicity

2.1 REGIONAL GEOLOGY

Don Edwards San Francisco Bay National Wildlife Refuge lies within the geologic province known as the Coast Range. Small mountain ranges and ridges that run parallel with the coast characterize the

Coast Range. The ridges are steep and wooded and the valleys are broad and flat. Most of the long ridges and valleys are fault slices that moved horizontally along the San Andreas system of faults. These fault slices are still moving north. The Refuge is located between the San Andreas and Hayward faults with the Silver Creek fault closest to the site. These northwest trending faults are right lateral faults. The geology consists of Pleistocene marine sedimentary rocks and sediment known locally as Bay mud. The

Pleistocene sedimentary rocks consist mainly of marine sandstone. The bay mud consists of soft clay and loose silt and sand. San Francisco Bay floods a broad structural depression in the Coast Range where the Sacramento and San Joaquin Rivers meet to flow into the ocean through the Golden Gate.

Water flows in and out of the bay with the tides. The water is mostly brackish with about 2.8% dissolved salts, 15 percent less than normal seawater. It is estimated that about 8 million cubic yards of sediment wash into the bay every year.

2.2 SEISMIC DESIGN PARAMETERS

The active faults in the area, and particularly the nearby Hayward fault, create conditions where seismic design issues may have a significant impact on structure design. The Maximum Magnitude (Mmax) for the Hayward Fault, located approximately 8 kilometers south of the project site, is estimated to be a 6.8 event. The Silver Creek fault located approximately 3 kilometers away is closer to the site, but has a lower Mmax associated event (Mmax = 6.1). The San Andreas Fault (Peninsula) is further from the site at approximately 22 kilometers. The American Association of State Highway and Transportation Officials

(AASHTO) criteria for seismic design of ordinary bridges, allows for damage to occur when the structure is subjected to 1000-year strong ground motion, but does not allow for collapse of one or more spans and loss of life. Bridges may suffer damage that requires repair or partial replacement. AASHTO, however, requires bridges to withstand small, more frequent earthquakes without damage. For the

Extreme Event I Limit State Design, an earthquake load and resistance factor of 1.00 should be used during the LRFD design of the bridge structure.

Recommended seismic response parameters for the La Riviere Marsh Bridge, Don Edwards San

Francisco Bay NWR Project design are based on the AASHTO-LRFD Bridge Design Specifications, 5th edition, 2010, and represents horizontal peak ground acceleration (PGA) with 7 percent probability of exceedance in 75 years (approximately 1000-year return period). The 1000-year return period uniform hazard spectrum for the La Riviere Marsh bridge site, located at 37.521-degrees latitude and -122.053-degrees longitudinal, was obtained in accordance with the AASHTO ground motion maps for the probabilistic horizontal acceleration values corresponding to specific peak ground acceleration (PGA) and the spectral coefficients, namely the short- and long- period ground acceleration (Ss and S1 respectively) and corrected for the soil profile at the bridge site.

Ground motions caused by earthquakes are influenced not only by the distance from the fault planes, but also by the geology and soils found at the site. Subsurface profiles similar to the ones at bridge site, with clay and sand soil layers that significantly differ in stiffness and density, will have amplified ground motions and a resonant period governed by the layer thickness and shear wave velocity of the materials.

The damage potential of strong ground motions with respect to a structure is typically affected by the period of the strong earthquake motion and the resonant period of both the soil and the structure. Based on the subsurface profile at the bridge site, the weighted average standard penetration test (SPT) N-values for the top 100 feet of subsurface materials was estimated between 15 and 50 with an average plasticity Index (PI) of 14. Therefore, the site is classified as Class D according to the site class definitions specified in Table 3.10.3.1-1 of ASSHTO-LRFD Bridge Design Specifications 5 th

Edition, 2010 (AASHTO-LRFD). Site Class D is defined as stiff soil with average shear wave velocity for the upper 100 feet of the soil profile between 600-feet per second and 1,200- feet per second, or with either average Standard Penetration Test (SPT) blow count for the upper 100 feet of the soil profile between 15 and 50 blows per foot, or average undrained shear strength for the upper 100 feet of the soil profile between 1,000 and 2,000 pounds per square foot.

A seismic hazard analysis to establish ground motions for seismic design was conducted. The recommended spectral acceleration coefficient values for probabilistic design with a return period of

1000 years were calculated using the program provided with the AASHTO-LRFD Bridge Design

Manual developed by the USGS (2008) entitled “Seismic Design Parameters”, version 2.10 and are summarized in Table 1.

TABLE 1: - Summary of Seismic Parameters Corrected for Site Class D

Horizontal Peak Ground Acceleration, (As) 0.600g

Horizontal Response Spectral Acceleration at Period of 0.2 sec, (SDs) 1.455g

Horizontal Response Spectral Acceleration at Period of 1.0 sec, (SD1) 0.876g

Site Factor at Zero-Period of Acceleration Spectrum, (Fpga) 1.00

Site Factor at Short-Period Range of Acceleration Spectrum, (Fa) 1.00

Site Factor at Long-Period Range of Acceleration Spectrum, (Fv) 1.50

The five percent damped design response spectrum for Site Class D is shown below in the Figure below and the supporting data is presented in Table 2.

FIGURE A - Design Response Spectrum Corrected for Site Class D

TABLE 2: - Summary of Design Response Spectrum Data for Site Class D

PERIOD (sec) Sa (g) Sd (in) COMMENTS

0.00 0.600 0.000 (PGA)(Fpga) = As

0.12 1.455 0.206

0.20 1.455 0.568 (Ss)(Fa) = SDs

0.60 1.455 5.155

0.70 1.252 5.991

0.80 1.095 6.847

1.00 0.876 8.559 (S1)(Fv) = SD1

1.20 0.730 10.271

1.40 0.626 11.983

1.60 0.548 13.694

1.80 0.487 15.406

2.00 0.438 17.118

2.20 0.398 20.542

2.40 0.365 22.253

2.60 0.337 23.965

2.80 0.313 25.677

3.00 0.292 27.389

Based on the long acceleration coeffi9cient SD1 value of 0.876g calculated as Fv.S1, the bridge site is assigned to seismic hazard Zone 4 according to Table 3.10.6-1 in AASHTO-LRFD Bridge Design

Specifications 5 th

Edition, 2010. Seismic hazard zones reflect the variation in seismic risk in different regions needing different requirements for design as depicted in Table 4.7.4.3.1-1 in AASHTO-LRFD.

0.2

0.4

0.6

0.8

1.2

1.4

1.6

0 0.5 1 1.5 2 2.5 3

S p ec tr a l

A cc el er a ti o n

S a g

Period, T (sec.)

With a site corrected PGA (As) as high as 0.600g, and a water table close to the ground surface most soil types, at most densities, are theoretically liquefiable. Consequently, liquefaction hazard maps for the area (CGS or the Association of Bay Area Governments) show moderate hazard where dense gravelly sand is not at or near the surface. This general area, near the east abutment of the Dumbarton Bridge, was monitored for surface expressions of liquefaction following the 1989 Loma Prieta earthquake, which generated a PGA of 0.20 to 0.30g in the area, and none were observed. Additionally, it is understood that the existing La Riviere Marsh Bridge, which has been in place for several decades and has been subjected to earthquakes, has not experienced seismic or liquefaction damage. These observations suggest that liquefaction has not been significant at lower PGA values.

The uniform hazard spectrum presented above is most commonly used for seismic design of ordinary structures and is developed based on representative earthquake magnitudes for the region. In high seismic zone areas such as this project, a detailed evaluation of individual magnitude and distance of earthquakes most contributing to the seismic hazard spectrum known as "deaggregation" may be performed for each spectral period on the uniform hazard spectrum. Deaggregation was performed for this project to determine the nearest source event, which was confirmed by the Caltrans ARS online application. Deaggregation hazard data for the PGA for a 5% in 50 years (975 year return period) based on the USGS 2002 records was obtained from https://geohazards.usgs.gov/deaggint/2002/.

https://geohazards.usgs.gov/deaggint/2002/

SECTION THREE–Subsurface Investigation

3.1 FIELD EXPLORATION

During the first phase of investigation, two borings (02B-1 and 02B-2) were advanced at the proposed replacement bridge abutment locations for sampling, to identify soil types, and to determine SPT N-values at various depths. Boring 02B-1 and 02B-2 were located near the proposed Abutments 1 and 2 respectively. A cone penetrometer test (CPT) sounding was conducted at each of two proposed pier locations (CPT-1 and CPT-2 near piers 2 and 1, respectively). Taber Consultants, of West Sacramento, California, conducted the drilling and CPT testing. The borings were accessed and drilled with a

Diedrich D-120 truck-mounted drill rig and were advanced using 100-mm inside-diameter hollow stem augers. Standard Penetration Tests (SPT’s) were taken at 150-mm intervals in accordance with

AASHTO T-206 using the "cathead and rope" technique with drill efficiency of 60 percent. Several undisturbed, thin-wall tube (Shelby) samples were also retrieved from the boring at various depths to evaluate the strength and settlement properties of the fine-grained soils encountered. The samples were retrieved and transported for laboratory testing in accordance with AASHTO T-207 using a 75-mm diameter Shelby tube.

Borings were located and logged by CFL personnel. Samples retrieved using the Standard Penetration

Test Sampler were tested at the CFL’s materials laboratory for gradation/classification, density, and

Atterberg limits. The Shelby tube samples were tested at Sierra Testing Laboratory, El Dorado Hills, California. Individual boring logs and photographs of the boring and CPT locations, and a subsurface profile are provided in the Appendix B of this report. A plan sheet with a summary of the boring logs with respect to the bridge foundations is also attached in Appendix B.

The cone penetrometer tests involve hydraulically pushing a cylindrical probe vertically through the soil at a constant rate to obtain continuous measuring of the resistance at the conical tip of the probe and along the steel shaft, and the pore-water pressure. The cone penetrometer probe used for the soundings is

36-mm in diameter, with 60-degree apex at the tip, 1000-mm projected tip area, and a 0.015-m sleeve surface area. No subsurface materials are retrieved from this test. All CPT testing was conducted in accordance with ASTM D-5778. The CPT results were used to estimate the soil types and stratigraphy of the subsurface using friction ratios, tip resistance and skin friction at 50-mm intervals along the tested column. The CPT test results obtained from CPT-1 and CPT-2 soundings, and interpretation of the results for characterizing soil types, evaluating liquefaction potential, and obtaining pile ultimate capacities were provided by VBI In-Situ Inc., a Division of Taber Consultants and are presented in

Appendix C.

The initial investigation identified a theoretically liquefiable stratum at depth but did not penetrate the subsurface deep enough for a thorough characterization of the unit. Also, there was some variability between what was logged from the drilling operation and what was interpreted from the CPT soundings.

To clarify this variability, calibrate the CPT to the SPT logging, and to explore greater depths, a second phase of investigation was conducted. Due to noise level limitations associated with hollow-stem auger drilling and potential delays in the project design, a second investigation of CPT only was executed later in 2003. CPT soundings were conducted near each of the abutment borings from the initial investigation and the soundings were taken to refusal. Analysis of the new soundings (CPT-1 FHWA near Abutment

2 and CPT-2 FHWA near Abutment 1) were also conducted by VBI In-Situ Inc., and results and interpretations are also presented in Appendix C.

3.2 SUMMARY OF SUBSURFACE EXPLORATION

Boring 02B-01, STA 10+439, 4.5 meters RT is located at the proposed Abutment 1 location for the La

Riviere Marsh Bridge. Boring 02B-01 encountered 0.2 meters of asphalt pavement and base course overlying red brown clay from 0.2 to 1.5 meters, brown clayey sand and silty sand from 1.5 to 3.0 meters, gray clay from 3.0 to15.0 meters, and brown silty sand from 15.0 to 15.45 meters. Boring 02B-

02, STA 10+476, 5.0 meters RT is located at the proposed Abutment 2 location. Boring 02B-02 encountered 0.2 meters of asphalt pavement and base course overlying brown clayey sand from 0.2 to

1.95 meters, gray clay from 1.95 to 16.0 meters, and brown silty sand from16.0 to 16.95 meters.

Samples were obtained for laboratory testing and classified as CL, CL-ML, and SM in the Unified Soil

Classification System (USCS). Table 3 summarizes the laboratory results of the tests performed on selected samples.

TABLE 3: - Summary of Laboratory Results from Borings at La Riviere Marsh Bridge

BORING

NUMBER

DEPTH

(m)

USCS % SAND

%FINES

<200

% MOISTURE PI LL

DRY

DENSITY

kN/m3

02B-01

4.5 – 4.95 CL 11 89 26.6 15 38 -

15.0 – 15.45 SM 69 16 15.4 NP NV -

02B-02

6.30 – 6.50 CL 8 92 25.3 17 39 16.0

7.50 – 7.95 CL-ML 46 52 26.9 7 23

10.77 – 10.92 CL 3 97 24.7 19 44 15.8

10.92 – 11.07 CL 1 99 24.0 20 47 16.3

15.25 – 15.40 CL 21 76 25.4 17 35 16.1

One dimensional consolidation testing was also performed on a Shelby Tube sample of low plasticity clay from 9.15-meters depth in boring 02B-02, and a series of three isotropically-consolidated undrained triaxial tests with pore-water pressure measurement was conducted on specimens from the same sample tube. The sample was chosen for testing based on the quality of the sample as well as the visual observations of the most conservative, higher plasticity clay logged during drilling. The results of the consolidation test indicate a preconsolidation pressure of 197 kilopascals (4,100 pounds per square foot;

however, sample disturbance may have impacted this interpretation), a recompression index of 0.03, and a compression index of 0.25). The coefficient of consolidation was determined to be 0.03 millimeters square per second (0.003 inches square per minute). The results of the triaxial text indicate undrained shear strength of 111 to 141 kilopascals at consolidation pressures of 144 to 258 kilopascals. The soil behavior is generally consistent with normal to slightly overconsolidated conditions, and the effective friction angle and cohesion are 23 degrees and 52 kilopascals, respectively. The consolidation and triaxial test results are presented in Appendix D.

CPT-1and CPT-2 soundings at the proposed pier locations were conducted with an electronic push cone.

The CPT soundings were located on the existing bridge beginning after coring through the existing bridge deck. CPT-1 sounding was located at STA 10+464, 5 meters RT. The following soil types were classified from CPT-1 results: 4.3 meters of clay overlying sandy silt to silty sand from 4.3 to 5.2 meters, silty clay to clayey silt from 5.2 to 10.0 meters, sandy silt to silty sand from 10.0 to 11.1 meters, silty clay to clayey silt from 11.1 to 13.1 meters, and sandy silt to silty sand from 13.1 to 13.4 meters.

CPT-2 sounding was located at STA 10+449, 4 meters RT. The following soil types were classified from CPT-2 results: 1.0 meter of clay overlying clayey silt and sandy silt from 1.0 to 1.5 meters, clay from 1.5 to 4.3 meters, sandy silt and silty sand from 4.3 to 6.7 meters, clay and silty clay from 6.7 to

11.6 meters, sand to silty sand from 11.6 to 13.7 meters, clay and silty clay from 13.7 to 14.9 meters, and silty sand from 14.9 to 16.5 meters.

CPT-1 West Side (also referred to as CPT-1-FHWA) and CPT-2 East Side (also referred to as CPT-2-

FHWA) were located near the proposed abutment locations. CPT-1 West Side is at approximate Station

10+477.5, 5 meters right of centerline, and CPT-2 East Side is at approximate Station 10+437.5, 4.5 meters right of centerline. The following soil types were classified from CPT-1 West Side soundings: the top 1.8 meters were classified as sand with silty sand and sandy silt lenses, sandy silt and clayey silt from 1.8 to 2.4 meters clay from 2.4 to 7.3 meters, clayey silt to silty clay from 7.3 to 8.2 meters, clay from 8.2 to 12.2 meters clayey silt to silty clay from 12.2 to 15.5 meters, sandy silt to clayey silt from

15.5 to 17.5 meters and sand to gravelly sand below 17.7 meters. The following soil types were classified from CPT-2 East Side soundings: the top 0.8 meters consists mainly of silty sand to sandy silt materials, clay to some silty clay from 0.8 to 7.6 meters with occasional thin sand and silty sand lenses, silty sand to sandy silt from 7.6 to 9.7 meters, clay from 9.7 to 15.0 meters, sand from 15.0 to 16.5 meters, clay with thin clayey silt lenses from 16.5 to 19.5 meters, and sand to gravelly sand below 19.2 meters. Both CPT soundings reached refusal at the bottom of the CPT tests. CPT-1 West Side advanced

17.7 m and CPT-2 East side advanced 20.2 meters prior to reaching refusal and strata below these depths were not penetrated by the CPT. The method of CPT interpretations is shown with the CPT results in Appendix C.

SECTION FOUR – Analysis and Recommendations

4.1 SITE CHARACTERIZATION

4.1.1 GEOTECHNICAL PROFILE

The investigations suggest that the subsurface can be characterized for design purposes as five essentially horizontal strata overlying a more competent, sixth stratum that was hard and therefore not easily penetrated by the CPT nor sampled by the SPT in the borings. This idealized stratigraphy is summarized in Figure 1 attached in the back of the report and in Table 4 below. The soils are predominantly low plasticity clay and silty clay, and are soft near the surface (Layer 1) and increase in stiffness from stiff to very stiff at depth (Layers 2 and 4). A variable-thickness lens (0 to 2.4 meters) of less plastic and more silty and sandy material is present in most of the explorations within the stiff to very stiff clay, and is identified as Layer 3. An attempt to sample from within this interpreted layer

(Layer 3) with a Shelby tube was unsuccessful, and interpretation is based on logging of cuttings and

CPT soundings. Layer 5 is the most sandy of the strata (16 percent) and is different from the overlying soil in that it is brown, rather than gray. Layer 5 was not fully penetrated with the borings but it was penetrated with the CPT soundings at each abutment. The two CPT soundings that penetrated Layer 5 reached refusal abruptly in Layer 6. The CPT interpretation of Layer 6 is that of gravelly sand that is much denser and harder than Layer 5. However, given the proximity of the bedrock outcrop and the proximity to an area of mapped moderate liquefaction hazard in local (California Geological Survey, Seismic Hazard Zone Map, Newark 7.5 minute quadrangle) and regional (http://gis.abag.ca.gov) maps, Layer 6 may be characterized as the top of weathered sedimentary or very dense gravelly sand layer.

The change in character between Layer 4 and Layer 5, and between Layer 5 and Layer 6, in conjunction with observations of the site setting, indicates, potentially liquefiable soils are not present below Layer

5. Based on this, it is anticipated that Layer 6 will be used for tip resistance and the dynamic test pile program should confirm the nominal resistance of this layer. This can be done during construction through dynamic testing and signal matching evaluation, as discussed in a subsequent section.

4.1.2 SCOUR

From a design perspective, conditions are fairly uniform across the site; however, separate analyses were performed for the abutments and piers to consider different pile top elevations and, possibly, different scour scenarios. Up to 2 meters of design scour has been estimated by the hydraulics section for this site, and is recommended for design. For the analyses presented here, it has been assumed that this scour would occur only at the pier locations and, with respect to elevation, would occur from 0.5 to -1.5 meters. Thus, -1.5 meter is the elevation below which pile axial resistance is developed at the pier locations. The bridge abutments are protected from this scour and that the top of the pile is at elevation

2.0 meters, approximately 2.5 meters below existing grade. If scour is to be designed for at the abutments, then the analyses for the pier locations are appropriate for use at the abutments.

http://gis.abag.ca.gov/

TABLE 4:- Soil Characterization

SOIL LAYERS

LAYER 1 LAYER 2 LAYER 3 LAYER 4 LAYER 5 LAYER 6

Top Elevation (m) Ground surface

(4.9)

2.0 -2.7 -5.1 -10.1 -14.5

Bottom Elevation (m)

2.0 -2.7 -5.1 -10.1 -14.5 Unknown

Thickness (m)

0.0 – 2.9 4.7 2.4 5.0 4.4 Unknown

Material Description (1) Silty clay and other

Clay and silty clay

Sandy silt, clayey silt, silty sand

Clay and silty clay

Mediums dense silty sand

Dense gravelly sand(2)

STATIC (PRE-LIQUIFACTION) PARAMETERS

Undrained Strength, su (kPa) 20 70 N/A 90 N/A N/A

Friction angle (deg.) N/A N/A 30 N/A 32 38

Unit weight, (3) /’

(kN/ m3)

17.5 / 7.0 16.8/7.0 16.8/7.0 16.8/7.0 17.5/7.7 20.0/10.2

LPile PLUS 3.0 soil type Soft Clay Stiff Clay

Medium

Submerged

Sand

Stiff Clay

Medium

Submerged

Sand

Medium

Submerged

Sand

Coef. of subgrade reaction, k, (kPa/m) 12,000 140,000 16,000 140,000 16,000 34,000

Strain @ 50%, 50 0.02 0.01 N/A 0.01 N/A N/A

Strain @ 100%, 100 0.1 0.05 N/A 0.05 N/A N/A

SEISMIC (POST-LIQUEFACTION) PARAMETERS

Undrained strength, su (kPa) 15(4)(8) 55(4)(8) N/A(6) 90(6) N/A(6) N/A(6)

Coef. of subgrade reaction, k, (kPa/m) 10,000(4)(8) 112,000(4)(8) 0(7)(8) 140,000(6) 16,000(6) 34,000(6)

Strain @ 50%, 50 0.02 0.01 N/A 0.01 N/A N/A(6)

Strain @ 100%, 100 0.1 0.05 N/A 0.05 N/A N/A(6)

1. Review boring logs for further information on material types.

2. Based on CPT record and refusal only, drilling and sampling did not penetrate this layer.

3. Water is at approximately 3.5 m elevation at the abutments and the ground surface elevation at the piers.

4. Based on an estimate of an approximately 20% reduction in strength following strong seismic shaking.

5. Estimated undrained residual shear strength of liquefied soil.

6. No liquefaction or softening so parameters remains the same as for static.

7. Assume zero strength in layer 3 following liquefaction.

8. Post-liquefaction strength above liquefiable layer 3 was neglected for lateral resistance calculations.

4.2 LIQUEFACTION POTENTIAL

Soil liquefaction is a phenomenon in which a cohesionless soil deposit below the groundwater table loses a substantial amount of strength due to pore pressure generation resulting from earthquake strong ground shaking. The reason for the generation of pore pressure is that cohesionless soils tend to compact during earthquake cyclical shaking and this tendency causes the pore water pressures in the solid to increase until the pore water dissipates from the soil layers. This pore pressure increase, in turn, causes a reduction in effective stress and associated reductions in soil strength and stiffness. Recently deposited and relatively loose natural soils and un-compacted or poorly compacted fill soils are susceptible to liquefaction. Submerged loose sand and silty sands are particularly susceptible to liquefaction. Loose silts and gravels may also have a potential for liquefaction. Dense natural, well-compacted fills and cohesive soils are known to have a relatively low susceptibility to liquefaction.

4.2.1 LIQUEFACTION POTENTIAL

Initial screening of site conditions for the potential for liquefaction was conducted to determine if a more detailed evaluation of liquefaction would be necessary. An evaluation for the potential for, and consequences of, liquefaction in soils near the surface were made in accordance with the following criteria (FHWA, 2011):

Seismic Hazard Level of the bridge, which depend on the value of Site Class-adjusted spectral acceleration at both long period (SD1) and short period (SDS).

Geologic age and depositional environment are indicators of liquefiable materials. Relatively young deposits (Holocene <11,000 years) deposited in potentially loose, uncompacted environments are susceptible to liquefaction.

Site history can indicate if any liquefaction has occurred in the past. According to past site survey’s following recent seismic events, there are no indications that liquefaction has previously occurred at the site (or in the vicinity of the site in similar geotechnical conditions). Even though records indicate that the existing bridge has been subjected to a Magnitude 6 earthquake and has been in place for over 34 years, it shows no signs of damage due to settlement that would be caused by liquefaction.

If soils are below the groundwater table, they may be susceptible to liquefaction.

Clayey soils with low sensitivity have a potential for liquefaction if one or more of the following criteria are met:

• Clay content (grain size < 0.005 mm) less than 15 percent;

• Liquid limit less than 35 percent; or,

• Natural water content greater than 90 percent of the liquid limit.

Clayey soils with highly sensitivity should not be screened out. Sensitive clays possess all of the following properties:

• Liquid limit less than 40 percent.

• Water content greater than 0.9 times the liquid limit.

• Liquidity index greater than 0.6

• (N1)60 less than five or normalized cone penetration resistance qc1, less than 20 ksf.

Sandy soils have a potential for liquefaction if the minimum corrected Standard Penetration Test resistance, (N1)60 values are less than 30 blows per foot or corrected cone penetration test tip resistance qc1, is less than 160 ksf with a sufficient number of tests.

The initial screening of liquefaction potential for the bridge site is summarized in Table 5 below, with each criterion related to the previously reported six soil layers in the design soil profile.

Preliminary analysis for liquefaction based on the criteria described above and detailed in Table 5 indicates soil layers 3 is potentially liquefiable and will require further evaluation of the potential impacts to the pile foundation. Soil layer 5 met the SPT criteria for potential liquefaction based on blows per foot below 30 (medium dense sand), but not the CPT criteria (dense sand). Since CPT data is taken continuously through the soil profile versus SPT data that is taken from singular point locations during drilling, the CPT data appears to provide more accurate information for layer 3. Also, liquefaction studies have shown a reduced possibility of liquefaction in layers exist at greater than 15 meters in depth primarily due to increased density and layer confinement at depth. Since layer 5 exists at greater than 15 meters below the ground surface, CPT data indicates dense sand rather than medium to loose sand, and the site history does not show past liquefaction, layer 5 is considered non-liquefiable.

Table 5: Initial Screening of Liquefaction Potential

Layer 1 Layer 2 Layer 3 Layer 4 Layer 5 Layer 6

Top Elevation (m) 4.9 2.0 -2.7 -5.1 -10.1 -14.5

Bottom Elevation (m) 2.0 -2.7 -5.1 -10.1 -14.5 Unknown

Thickness (m) 2.9 4.7 2.4 5.0 4.4 Unknown

PHYSICAL PROPERTIES

Material Description Silty clay and other

(SM/SC)(1)

Clay and silty clay

(CL)(3)

Sandy silt, clayey silt, silty sand (CL-

ML)

Clay and silty clay

(CL) (1)(3)

Meduim

Dense Silty sand (SM)

Dense Gravelly sand (GM/SM)

Natural Moisture

Content (%)

- 26.6 26.9 25.4 15.4 -

Liquid Limit (%) - 38 23 35 NV -

Plasticity Index (%) - 15 7 17 NP -

Percent Fines (%) - 89 52 76 16 -

DENSITY

SPT N1(60)

(4) 4 12 8 14 20 -

CPT qC1

(4) 67 - 89 - 163 488 (refusal)

WATER TABLE

Yes Yes Yes Yes Yes Yes

SCOUR

Full Layer Partial Layer

(Elev. -

1.5m)

No No No No

SEISMIC HAZARD LEVEL

(2)

IV IV IV IV IV IV

GEOLOGIC SETTING

Pleistocene

Marine

Sedimentary

Pleistocene

Marine

Sedimentary

Pleistocene

Marine

Sedimentary

Pleistocene

Marine

Sedimentar y

Pleistocene

Marine

Sedimentary

Pleistocene

Marine

Sedimentary

Layer 1 Layer 2 Layer 3 Layer 4 Layer 5 Layer 6

SITE HISTORY

No liquefaction observed

No liquefaction observed

No liquefaction observed No liquefaction observed

No liquefaction observed

No liquefaction observed

POTENTIAL FOR LIQUEFACTION

Yes No Yes No Borderline No

Key Properties and

Comments on liquefaction potential

This layer is neglected for scour regardless of liquefaction

Not liquefiable:

LL>35%

Clay

%>15%

Potentially liquefiable:

SPT N1(60)<30

(loose sand) and CPT qC1<160ksf

(loose sand)

Not liquefiable:

LL>35%

Clay

%>15%

Not liquefiable:

SPT

N1(60)<30

(loose sand), But CPT qC1>160ksf

(dense sand)

Not liquefiable:

Correlated SPT

N1(60)>30

(dense)

1. Design parameters presented. When multiple test results available for one unit, then the most conservative value was used as the design parameter.

2. Based on FHWA (2011) “LRFD Seismic Analysis and Design of Transportation Geotechnical Features and Structural Foundations”, GEC No. 3.

(Rated I, II, III, and IV with IV being the highest level)

3. No sensitive clays were observed on site according to FHWA (2011) criteria (LL<40%; mc>.9xLL; liquidity index>0.6; and SPT N1(60)<5 or CPT qC1<20ksf).

4. Density values for SPT and CPT data were averaged for each layer.

4.2.2 LIQUEFACTION INDUCED SETTLEMENT

A consequence of earthquake induced liquefaction is the subsequent post-earthquake dissipation of excess pore-water pressure, potentially resulting in soil settlement and downdrag loads. Soil layer 3 is considered liquefiable and was further analyzed for post-liquefaction induced settlement. Based on experimental studies, Tokimatsu and Seed (1987) developed a simple chart based procedure for estimating post liquefaction settlement. The Tokimatsu and Seed (1987) chart for a magnitude 7.5 earthquake plots the volumetric strains corresponding to the blow counts (N160) and cyclic stress ratios

(CSR) for each liquefied layer should be multiplied by the layer thickness to estimate settlement.

Following this method (as recommended in FHWA, 2011), maximum soil layer 3 settlement value was estimated at approximately 3 inches (76 mm). The calculation Table and Charts are attached in

Appendix E.

4.2.3 PRE- AND POST-LIQUEFACTION PROPERTIES FOR LATERAL RESISTANCES

Lateral pile resistance for static and seismic loading at the Service and Extreme Event I Limit States should be evaluated by the structural engineer using the parameters in Table 4. These material strength values were estimated based on recommendations published by Seed and Harder (NCHRP 12-70). These investigators found little evidence of cyclic strength loss in cohesive soils during earthquakes and noted that cohesive soils also tend to show an increase in undrained shear strength when subject to rapid cyclical loading. They concluded that for earthquakes of greater than magnitude 7.0, a strength loss of

10-15 percent is recommended. Also Makdisi and Seed (1978) found that substantial permanent strains may be produced by cyclic loading of clay soils to stresses near the yield stress, while essentially elastic behavior is observed for large numbers of cycles of loading at cyclic shear stresses of up to 80 percent of the untrained strength. These researchers recommend the use of 80 percent of the undrained strength as the “dynamic yield strength” for soils that exhibit small increases in pore pressure during cyclic loading, such as clayey materials.

Typically, based on tests by Seed and Others, saturated sands that liquefy possess some “residual” shear strength even when in the liquefied state. In initially loose soils, the residual strength may be very small and of little consequence. The seismic (post-liquefaction) parameters in Table 4 assume a 20 percent reduction in strength and stiffness for clayey soil layers and 90 percent for silty and sandy layers.

Because Layer 1 is within the scour zone, we recommend that this layer will not be accounted for in the axial and lateral foundation design calculations. Based on calculations presented in this memorandum it appears that Layer 3 will liquefy, and therefore layers below the liquefiable zone (Layers 4 and 5) can be incorporated in the lateral foundation design. Also due to the aforementioned research references and the fact that the existing bridge is founded on shallow spread footings and has performed well through at least one large earthquake it is recommended that the reduced residual soil strength values listed in

Table 4 for the lateral movement calculations during an earthquake are used in the design.

4.2.4 LATERAL SPREAD DISPLACEMENT

Lateral spreading following an earthquake may occur beneath a bridge approach or abutment fills if the under laying soil is considered liquefiable. The static factor of safety of the soil mass is near the bridge abutment is greater than one but when lateral movements caused by an earthquake induce shear stresses that exceed the soil shear strength of the liquefied soil, lateral spreading may occur. Lateral spreading is not anticipated at the bridge site due to the depth of the potentially liquefiable layer providing confinement of the potentially displaced mass.

4.3 BRIDGE FOUNDATION DESIGN RECOMMENDATIONS

Due to the foundation soil conditions and the close proximity to dense gravelly sand, precast pre-stressed concrete driven piles are recommended for the bridge foundation. Pre-stressed concrete piles are recommended because of their resistance to corrosion, cost, availability, and high load bearing resistance. Although corrosion tests were not conducted, the soils in this area are considered corrosive due to both the depositional environment of the materials and the current marine environment of the site.

Square precast (356-, 406-, and 457- millimeters) concrete piles consistent with standard Caltrans pile sizes were valuated as foundation elements for bridge piers and abutments. The recommendation for the pile is presented in Table 6.

4.3.1 NOMINAL COMPRESSION RESISTANCE

The static compressional pile resistance was mainly estimated using the program Driven (version 1.2).

Taber Consultants also used the program CPTINT to estimate the pile resistance using CPT for comparison. The Driven program is based on soil parameters data obtained from SPT N-values for cohesionless soils and undrained shear strength for cohesive soils. The nominal compression resistance should be established by adding the side resistance to the tip resistance. The idealized soil profiles shown in Figure 1 were used in the analyses at the pier and abutment locations and soil parameters included in Table 4. Analyses were conducted for both the piers and abutment using the aforementioned three reinforced concrete pile sizes (356-, 406-, and 457-millimeters concrete piles). Figures 2 through 7 show the nominal (unfactored) resistance for different pile depths from the Driven analysis at the proposed pier and abutment locations.

The CPTINT software uses data from the CPT soundings to estimate pile nominal (unfactored) resistance (See Appendix C). CPTINT analysis was conducted only for the 356-mm square reinforced concrete pile based on the Nottingham and Schmertmann Method. A comparison of the two methods of calculating nominal axial resistance of 356-millimieter pile (Driven vs. CPTINT) is presented in Figures

8 and 9. The Driven results are in general agreement with the CPTINT results and it is recommended

Driven be used for design. Although the larger diameter pile (457 millimeters) was selected for this bridge, the comparison between both resistance calculation methods was verified. If treatment of scour is different than assumed herein (Section 4.1.2), the resistance indicated in the figures should be revised by adding or subtracting resistance, as appropriate, from the revised scour depth. The resistance available to resist downdrag loads at the Extreme Event I limit State is limited to the lesser of the structural resistance, or the geotechnical resistance developed below elevation -5.1 meter. The skin, tip and total nominal resistance developed for the piers and abutments for the 457-millimeter pile recommended for the bridge foundation design are presented in Figures 6 and 7 respectively.

For the Extreme Event I Limit State, to mitigate the potential for post-liquefaction settlement of the bridge, the bearing resistance developed below the liquefaction should be greater than the sum of the design and the downdrag loads. The downdrag load was calculated as 200 and 450 kilonewtons for the piers and abutments respectively based on the potential settlement of the clay layers in addition to the potential strength loss in the liquefiable layer 3 (above -5 elevation) following an earthquake. This should be treated as an extreme event loading state and no further reduction in resistance should be applied (resistance factor of 1.0). In accordance to ASSHTO guidelines a load factor of 1.4 should be applied to the downdrag loads. If this available geotechnical resistance is exceeded, and the structural pile resistance is not, settlement of the pile should be expected following liquefaction. If the structural resistance is exceeded by the combination of downdrag and structure loads, lasting pile damage can be expected. The nominal pile resistance available to support the structure plus downdrag loads was estimated by considering the positive skin friction and tip resistance below the layer contributing to downdrag (layer 3).

To mitigate the liquefaction effects on stability, downdrag loads can be reduced by isolating the upper part of the pile from the soil with low friction material, or the geotechnical resistance can be increased by driving the piles deeper into Layer 6, if needed. The Driven results indicate nominal friction resistance develops at a rate of approximately 500 kilonewtons per meter for a 457-milimeter square pile in dense gravelly sand, so one meter of penetration in the dense gravelly sand layer will significantly increase the geotechnical resistance. For conservative estimation of available pile resistance for geotechnical design, it is assumed that the piles will meet refusal within 2 m in the dense gravelly sand layer.

The resistance factor recommended for Geotechnical Strength I Limit State in axially loaded piles is 0.7 and the resistance factor for the Service I and Extreme Event I Limit States is 1.0. These factors are in accordance with AASHTO-LRFD and are based on nominal resistance using the methods above and the methods of construction control described below. Because soil Layer 3 is susceptible to strength loss during strong ground shaking or liquefaction, the Extreme Event I Limit State should be evaluated for compressional loading and should account for a downdrag load within the top layers (above elevation-5) which is produced due to material strength loss due to temporary pore-water pressure build up in the soil.

A summary of the loading conditions and required pile resistances at the each Limit State for design are listed in Table 6. The loads were obtained from the structural engineer and the required factored nominal resistances were calculated using resistance factors of 0.7 for the Strength Limit State and 1.0 for both the Service and Extreme Event I limit States. The Strength Limit State condition controls the design of the piles, as indicated by the last column, denoting the nominal driving resistance required.

TABLE 6:- Foundation Recommendations for Each Abutment and Pier Piles

(1)

S u p p o r t

L o ca ti o n

P il e T y p e

C u t-o ff E le v m

S e r v ic e -I

L im it

S ta te L o a d k N

T ta l

P er m is si b le u p p o r t

S e tt le m e n t

(m m

Required Factored Nominal Resistance (kN)

D e si g n

T ip

E le v .(8

(m )9 p e c if ie d

T ip

E le v

(m

U n fa c to r e d m in a l

D ri v in g

R si st a n c

R q u ir e d

(3 k

Strength Limit(3) Extreme Event I Limit

Nominal

Resistance

(kN) Comp.

(Φ=0.7)

Tension(5)

(Φ=0.7)

Nominal

Resistance

(kN)

Comp.(6)

(Φ=1.0)

Tension(5)

(Φ=1.0)

Abuts

(7)

457-mm

Square Pile

2.8 to

3.0

1,341 25 2,769 1,938 Not available 1,691 1,691 0 -14.7 -14.7 2,769

Piers(7)

457-mm Square

Pile

2.8 to

3.6

1,435 25 3,034 2,124 Not available 1,661 1,661 0 -14.5 -14.5 3,034

1. Per pile values presented in accordance with Caltrans guidance in Memorandum to Designers 3-1 (Deep Foundations) dated July 2008.

2. Cut-off elevation changes for each pile due to super elevation (slope) on the bridge deck. See plans for specific elevations.

3. Assume use of embedded sensors for testing of pile integrity.

4. No setup and no relaxation assumed due to bearing on dense gravelly sand.

5. No uplift, so no required factored nominal resistance in tension in the strength limit state nor the extreme limit state.

6. Extreme event I limit state includes the factored downdrag load on the pile.

7. Abutment support consists of a single row of 5 piles and pier support consists of a single row of 7 piles.

8. Minimum tip elevations are based on compression demands only at the strength I limit state and extreme event I limit state requirements. Lateral load demands will be determined by the structural engineer and may control the design.

4.3.2 NOMINAL LATERAL PILE RESISTANCE

Lateral resistance can be evaluated using p-y curves to model the non-linear load deflection response of piles and pile groups. The deflection should be compared to acceptable limits and pile stresses should be compared to pile resistance. At the Strength I Limit State, the pile or pile group is subjected to the factored lateral and axial loads and factored moment, and the resulting factored axial and bending stresses are compared with the factored axial and bending resistances. A resistance factor of 1.0 should be used for the Extreme Event I Limit State and Service I Limit States for lateral resistance calculations.

A simplified soil stratigraphy suitable for this analysis, and representative soil parameters for static and seismic (post-liquefaction) loading, are provided in Table 4.

4.3.3 GROUP EFFECTS

A group of piles may have lower nominal axial resistance in clay than the sum of their individual resistances. For pile groups in clay, the nominal axial resistance of the pile group shall be taken as the lesser of the sum of the individual nominal resistances of each pile in the group, or the nominal resistance of an equivalent pier consisting of the piles and the block of soil within the area bounded by the piles. Piles should be spaced apart at a distance at least 3 times the pile diameter, center to center, to avoid axial loading group effects. If the cap is not in firm contact with the ground, the individual resistance of each pile shall be multiplied by an efficiency factor of 0.65 for a center-to-center spacing of

2.5 diameters or an efficiency factor of 1.0 for a center-to-center spacing of 6.0 diameters. For intermediate spacing’s, the efficiency factor may be determined by linear interpolation. If the cap is in firm contact with the ground, no reduction in the efficiency shall be required.

The P-y method of analysis should be used to evaluate the pile group horizontal movement. The values of P shall be multiplied by P-multiplier values, Pm, to account for group effects. The values of Pm are provided in the Table 7.

TABLE 7:- Pile P-Multipliers, Pm for Multiple Row Shading

Pile center-to-center spacing

(in the direction of loading)

P-Multipliers, Pm

Row 1 Row 2 Row 3 or higher

3B 0.7 0.5 0.35

4B 1.0 0.85 0.7

If the loading direction for a single row of piles is perpendicular to the row, a group reduction factor of less than 1.0 should only be used if the pile spacing is 5B or less. The deflection should be compared to acceptable limits and pile stresses should be compared to pile resistance.